C&ENVENG 4112/7112

Advanced Civil Geotechnical Engineering

Prof. Mark Jaksa

Dr. An Deng Brendan Scott

 

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THE UNIVERSITY OF ADELAIDE SCHOOL OF CIVIL, ENVIRONMENTAL AND MINING ENGINEERING

ADVANCED CIVIL GEOTECHNICAL ENGINEERING

(C&ENVENG 4112, 7112)

M. B. Jaksa

LECTURE SERIES No. 1

CHARACTERISTICS OF EXPANSIVE SOILS 1. INTRODUCTION This module examines the nature, behaviour and classification of expansive soils; the design of residential footings using a variety of engineering methods; and the assessment and remediation of structures damaged as a result of expansive soil movements. We first begin by exploring the nature, mineralogy and properties of expansive soils. 2. THE NATURE OF EXPANSIVE SOILS An expansive soil is one that exhibits some degree of volume increase, or swell, with the addition of moisture, or volume decrease, or shrinkage, with the removal of moisture. Expansive, or reactive, soils are a worldwide phenomenon, with their presence being observed in all but the polar continents. As shown in Figure 2.1, countries where expansive soils have caused damage include the United States, Australia, Canada, South Africa, India, Israel, Mexico, England, France, Spain, China, Myanmar, Peru, Brazil, Colombia, Venezuela, Cuba, Ethiopia, Ghana, Sudan, Saudi Arabia, Jordan, Iran, Morocco, Zimbabwe, Turkey.

Figure 2.1 Worldwide distribution of expansive soils (Chen 1975).

Regions where expansive soils are known to occur

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Expansive soil research has been led primarily by the United States, Australia, South Africa and Canada. The distributions of expansive soils in the United States and Australia are shown in Figures 2.2 and 2.3, respectively.

Figure 2.2 Distribution of expansive soils in the United States (Steinberg 1998).

Figure 2.3 Distribution of expansive soils in Australia (Steinberg 1998).

Expansive soils are of particular importance to geotechnical engineering in South Australia as many of Adelaide’s soils are expansive, as will be seen later. Problems associated with expansive soils were first recognised by geotechnical engineers in the latter part of the 1930s (Chen 1975). Expansive soils cause damage to residential and light commercial structures (Figure 2.4), as well as roads (Figure 2.5), airfields, other pavements (Figure 2.6), pipelines and canals. Damage caused by expansive soils exceeds the sum of all natural disasters and conservatively exceeds $10 billion in the US alone (Steinberg 1998).

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Figure 2.4 Wall cracking due to expansive soil movement.

Figure 2.5 Road undulation due to expansive soil movement.

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Figure 2.6 Surface cracking due to expansive soil movement. In order to understand why expansive soils behave the way they do, we need first to examine their structures and mineralogy. 3. THE MINERALOGY OF CLAYS Clay minerals are composed of essentially two main structural building blocks; the silicate tetrahedral unit, Figure 3.1, and the aluminium or magnesium octahedral unit, Figure 3.2. These basic units combine in various ways to form extensive sheets and are responsible for the general flat plate characteristics of clay particles, where two dimensions greatly exceed the third. The tetrahedral sheet is basically a combination of tetrahedral silica units which consist of four oxygen atoms at the corners, surrounding a single silicon atom. On the other hand, the octahedral sheet is basically a combination of octahedral units consisting of six oxygens or hydroxyls, depending on which is needed to balance the structure, enclosing an aluminium, magnesium, iron, or other atom. All clay minerals consist of these two basic sheets which are stacked in certain unique configurations and with certain cations present in the tetrahedral and octahedral sheets. Before we focus on some of the more common clay minerals, we need to examine the forces that hold these sheets together. 3.1 Inter-unit Bonding The forces that act between sheets, whilst being relatively weak in comparison to the primary bonds, that is the ionic and covalent bonds, are important sources of attraction between very small particles and between liquids and solid particles. Essentially, there are two types of secondary bonds; the hydrogen bond and the van der Waals bond.

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Figure 3.1 Silicate tetrahedral unit.

Figure 3.2 Aluminium or magnesium octahedral unit. The hydrogen bond is one formed by the attraction of oppositely charged ends of a permanent dipole, in this case, hydrogen as the positive, and usually oxygen as the negative. The hydrogen bond is considerably stronger than other types of secondary bonds and plays an important role in determining some of the clay mineral characteristics and in the interaction between soil particles and water. The van der Waals bond is similar to the hydrogen bond, but is caused by the attraction of instantaneous dipoles which arise from an imbalance of electrons in an atom. While hydrogen bonds are generally a magnitude stronger, their effect decreases more rapidly with distance than does the van der Waals bond.

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3.2 Kaolinite - Al2Si2O5(OH)4 The kaolinite crystal is made of repeating layers, each layer consisting of a silica-tetrahedral sheet and an alumina-octahedral sheet, sharing a common layer of oxygen atoms between them, as shown in Figure 3.3. Successive layers of the basic layer are held together by hydrogen bonds between the hydroxyls of the octahedral sheet and the oxygens of the tetrahedral sheet. Since the hydrogen bond is very strong, it prevents hydration and allows the layers to stack up to make a rather large crystal. A typical kaolinite crystal may be 70 – 100 layers thick, approximately 0.05 µm.

Figure 3.3 Atomic structure of kaolinite. Kaolinite is a weathering product of igneous and metamorphic rocks mainly by the alteration of feldspars usually by weathering in wet climates and under good drainage conditions, and is the chief clay mineral in most residual or transported clays. Soils with significant amounts of kaolinite are usually non-expansive (i.e. they do not exhibit a large degree of shrink or swell upon the addition or removal of water) and are often used in the pottery and ceramic industries. 3.3 Montmorillonite (Al,Mg,Fe3+)2(Si,Al)4O10(OH)2[Ca,Na,K] Montmorillonite, sometimes referred to as smectite, is made up of repeating three-sheet layers, two silica-tetrahedral units enclosing an alumina-octahedral unit, as shown in Figure 3.4. The bond between layers is by means of the van der Waals force which is relatively weak, and as a result, water enters easily because of a net negative charge deficiency in the octahedral sheet, causing swelling of the clay up to several times its own volume. When the layers become separated by water molecules several layers in thickness, they can be considered to act as individual particles which would be from one to five layers thick. Montmorillonites occur in sediments of semi-arid regions and are formed by degradation of some igneous rocks and volcanic ash under conditions of restricted drainage. Bentonite is a specific form of clay with a high montmorillonite content. Montmorillonites are extremely expansive and the swelling pressures developed can easily damage light structures and road pavements.

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Figure 3.4 Atomic structure of montmorillonite. 3.4 Illite The crystal structure of illite is the same as that for montmorillonite except that the repeating layers are bonded relatively weakly by potassium ions, as shown in Figure 3.5. Illite soils are derived from the weathering of acidic igneous and metamorphic rocks in a cool and moist environment under conditions of restricted drainage. Illites are relatively stable and, as a result, are fairly abundant.

Figure 3.5 Atomic structure of illite.

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3.5 Clay Minerals, Water and Expansion The expansiveness of a soil is related to the adsorbed water layer and the relative sizes of the clay minerals. Water is attracted to the surface of a clay mineral by means of three forces; the hydrogen bond, van der Waals force and also by the negatively charged clay surface which attracts cations present in the water. Figure 3.6, below, shows the relative sizes of the three clay minerals, discussed previously, when viewed from their edge. Figure 3.7 shows a montmorillonite and a kaolinite crystal with adsorbed water layers. The thickness of the water layers is approximately the same in each case but the ratio of thickness of clay particle to thickness of water layer is significantly greater for the montmorillonite crystal than for the kaolinite crystal. This phenomenon greatly influences the behaviour and the expansiveness of the clay minerals.

Figure 3.6 Relative sizes of clay minerals. Figure 3.7 Relative sizes of adsorbed water layers.

As a result of the relative sizes of the clay minerals and the adsorbed water layers, montmorillonite is more expansive than illite which, in turn, is more expansive than kaolinite. 4. SOIL TYPES OF THE ADELAIDE REGION As mentioned earlier, many of the soils found in the city of Adelaide are extremely expansive. The distribution of expansive soils in the Adelaide area is shown in Figure 4.1. Of particular note are the black earths, which are particularly expansive, with surface heaves of up to 200 mm, and the red-brown earths. These are described in some detail in this section. However, before doing so, it is necessary to examine briefly the master horizons adopted by soil scientists.

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Figure 4.1 Soils of the Adelaide metropolitan area.

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4.1 Soil Science Master Horizons Soil scientists divide a profile into a number of master horizons, each given a letter (O, A, E, B, C or R), as shown in Figure 4.2. The uppermost layer generally is an organic (O) horizon. It consists of fresh and decaying plant residue from such sources as leaves, needles, twigs, moss, lichens, and other organic material accumulations.

Figure 4.2 Soil horizons. The A horizon is a surface horizon composed of minerals and organic matter. Plant roots grow in this horizon. The organic matter is accumulated from growing plants and decomposed organic matter by organisms. Plants add organic matter to the A horizon as their roots grow. Organisms add organic matter to the A horizon as the decompose organic matter from the O horizon. For instance, earthworms decompose matter from O horizons, adding to A horizons, often drastically changing the soil profile. The E horizon is named for being the zone of eluviation, or leaching out. Minerals are translocated, or eluviated out of this zone, leaving mostly silica. Since the remaining silica is white, the E horizon may appear lighter in colour than its surrounding horizons.

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The B horizon is the zone of illuviation, or translocation in. It is usually located below an A, E, or O horizon. Clay, iron, humus or carbonates accumulate in B horizons, giving them more colour than surrounding horizons. This leaching in alters the character of the parent material. The C horizon is little affected by soil forming processes. It is characterised by lacking properties of the A, O, B or E horizons and of the parent material. The R horizon is the rock horizon, composed of hard bedrock. 4.2 Red-brown Soils As can be seen from Figure 4.1, red-brown (RB) earth soils comprise approximately 70% of the Adelaide metropolitan area. Taylor et al. (1974) identified 9 different red-brown earth profiles, RB1 to RB9. As shown in Table 4.1, the red-brown earths vary from moderate (M) to extreme (E) reactivity. These classifications will be discussed in greater detail later. Typical RB profiles are shown in Figure 4.3. Table 4.1 Reactivity of Adelaide soil profiles (Table D4, AS 2870).

4.3 Black Earths Black earth (BE) profiles are the most reactive soils in Adelaide and among the most reactive in the world. As shown in Figure 4.1, they are found in pockets along the foothills and especially in the Newton, Modbury, Holden Hill areas, where they have caused considerable damage to houses, roads and pavements. A typical BE profile, which is sometimes colloquially referred to as a ‘Bay of Biscay’ soil, is shown in Figure 4.4. As indicated in Table 4.1, the black earths are extremely reactive. Whilst the surficial, highly plastic, black clay is expansive, it is not as reactive as the clay that underlies it. Known as the

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(a) RB3 (b) RB5

Figure 4.3 Red-brown earth profiles (Taylor et al. 1974).

Figure 4.4 Typical black earth profile (Taylor et al. 1974).

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Keswick or Hindmarsh Clay, depending on the location, these clays are extremely reactive, as detailed below. 4.4 Keswick and Hindmarsh Clays Among the most reactive soils in the Adelaide area, and the world, are the Keswick and Hindmarsh Clays. As shown in Figures 4.5 and 4.6, these clays underlie most of Adelaide. They are highly plastic clays that have been overconsolidated as a result of desiccation. They are usually grey-green in colour, with red-brown or yellow mottling, and distinguished by slickensided joints and microfissures.

(a) (b)

Figure 4.5 Distribution of (a) Keswick and (b) Hindmarsh Clays (Sheard and Bowman 1996). These clays often underlie BE profiles, and the reactivity of the site is exacerbated when these clays occur closer to the ground surface. 4.5 Gilgais The term gilgai refers to dome-type undulations of the upper surface of the Keswick Clay and Hindmarsh Clays, as shown in Figures 4.7 and 4.8. The term gilgai derives from the Aboriginal word for small waterhole and refers to the tendency for surface water to pond in hollows between the undulations in wet weather. An abundance of gilgai in the Adelaide city area gives rise to marked lateral variation in soil type for the first few metres below the ground surface.

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Figure 4.7. Typical gilgai structures within the Adelaide city area (Selby and Lindsay 1982).

Figure 4.8. Examples of gilgai structures (Sheard and Bowman 1996).

Stapledon (1970) proposed a scenario for the development of gilgai structures, a summary of which is given below, and a graphical representation is shown in Figure 4.9. 1. Firstly, the Keswick and Hindmarsh Clays were deposited in a flocculated state in a river flood

plain.

2. Subsequently, the surface of the clay dried, cracked and became desiccated, largely as a result of uplift. This resulted in the formation of numerous vertical shrinkage cracks and the pseudo-consolidation of the clays, referred to earlier.

3. Wind-blown quartz sand and calcareous silt accumulated on the surface of the clay and penetrated and filled the cracks and open joints to form dykes.

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Figure 4.8 Suggested origin of gilgai structures (Stapledon 1970, Selby and Lindsay 1982).

4. Subsequent wetting, which led to the deposition of the red-brown Callabonna Clay, resulted in

the swelling of the Keswick and Hindmarsh Clays. 5. The pressures induced by the increased moisture resulted in upward swelling and doming of the

clays, as it was largely unconfined in the vertical direction. Horizontally, the clay was confined by the surrounding soil mass and lateral swelling resulted in the formation of the gently dipping joints as a consequence of shear failure.

The gilgai structures appear to be relatively recent in origin, as they have displaced soil-profile horizons (Selby and Lindsay 1982). It appears that most of the gilgais are inactive, though it is thought that they could be reactivated by a local increase in groundwater flow. 5. EXPANSIVE SOIL SAMPLING PRACTICE Expansive soil samples are obtained for assessment and testing, for the purposes of residential footing design, usually in a semi-disturbed state, using either the dynamic push or static push techniques. In most cases, the dynamic push technique is used. 5.1 Dynamic Push Technique The dynamic push technique is conducted using either a hand-held machine or truck-mounted rig. The hand-held device, as shown in Figure 5.1(a), is a jack hammer modified by replacing the chisel with a blunt rod that is placed inside a removable bush on top of the drilling rod. The hand-held device provides access to otherwise difficult-to-sample areas, as shown in Figure 5.1. The truck

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mounted rig (Figure 5.2), on the other hand, allows more efficient, deeper and less labour-intensive drilling.

(a) (b)

Figure 5.1 Dynamic push method (a) hand-held device, (b) truck mounted rig. In relation to the hand-held device, the drilling proceeds in the following manner: 1. Driving, from the ground surface, a hollow steel tube, generally 40 mm in diameter by 1.1 m,

1.6 m or 3 m long, into the soil by means of the modified jack hammer [Figure 5.2(a)];

2. The tube is then withdrawn from the ground by means of a tripod and chain block [Figure 5.2(b)];

3. The tube is then emptied into a core tray by upending it and gently tapping on the side of the tube, as shown in Figure 5.3.

4. Steps 1 – 3 are repeated until the desired depth is achieved, which is generally 3 m in Adelaide, and occasionally 4 m.

Truck-mounted dynamic push is performed in the same manner, however, the rods are typically 1 m in length, and the rods are extracted from the ground using the drilling rig’s hydraulic ram, as shown in Figure 5.4. 5.2 Static Push Technique The static push technique involves pushing a hollow tube into the ground in a fashion similar to that adopted when obtaining an undisturbed tube sample using a geotechnical drilling rig. Unlike the dynamic push method, which involves vibrating a tube into the ground, the static push technique using a consistent and relatively slow (approx. 20 mm/sec) push. In this way, a more undisturbed sample is obtained. Traditional drilling rigs, such as those shown in Figure 5.5, are often used.

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(a) (b)

Figure 5.2 Hand-held dynamic push method (a) driving the sampling rod into the ground, (b) removing the rod with a tripod and chain block.

Figure 5.3 Emptying soil into the core box. Figure 5.4 Truck-mounted dynamic push.

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Figure 5.5 Typical static push drilling rigs.

5.3 Placement in the Core Tray The core boxes in current use in Adelaide are 750 mm in length and they are generally placed in the trays as shown in Figure 5.6, or occasionally as shown in Figure 5.7.

Top

0.00

0.751.5

2.25

0.75 2.25

3.0

1.5

Bottom

BH1 BH2

Read  core

Depth  (m)

Figure 5.6 General method of placement of soil in the core box.

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BH1

BH2

Read core

BottomTop

0.75

1.52.253.0

0.751.52.25

0.00Depth (m)

Figure 5.7 Occasional method of placement of soil in the core box. 5.4 Aspects to be Mindful of When Sampling Expansive Soils If you are unsure which method has been used to place the soil in the core box, contact the driller

or, in some cases, obvious signs such as grass, topsoil, crushed rock, or asphaltic concrete, can suggest the ground surface and, hence, the orientation of the core within the box.

Particularly in stiff and moist, highly plastic soils, such as the Keswick or Hindmarsh Clays, the core sample has a tendency to grow in length or stretch, probably as a result of stress relief. Such stretch can be significant and will affect the interpretation of depths to layer boundaries and other stratigraphic features.

Be aware of the following aspects, as they are not part of the soil profile, but the sampling process, itself: o In some soil profiles, particularly sites where the profile goes from dry to wet, such as in

Prospect, oil (usually vegetable) is applied by the driller to the sampling tubes to prevent them from being locked into the underlying soil.

o In addition, sites which are underlain by stiff, highly-plastic clay, such as the Keswick and Hindmarsh Clays, the driller often pours dry sand down the sampling tubes to prevent the core samples from slipping out of the tubes during withdrawal.

Particularly if a borehole is being drilled on a site which is paved with asphaltic concrete, it is common for some of the pavement gravel to dislodge at the surface and fall into the borehole during sampling.

The presence of fill within a subsurface profile is a vital part of any geotechnical investigation. The logger should always be alert to any signs of fill present in the core sample. Indicators to the presence of fill include:

un-natural materials at depth, such as: glass, brick and concrete fragments; plastic; metal; timber;

mixtures of many different soil types over a small depth interval. e.g. clay, sand and gravel mixed in a seemingly random manner. One needs to be careful in categorically stating that such material is fill. (This material may be the remnants of an ancient alluvial deposit. However, if the gravels are relatively angular, an alluvial deposit is unlikely.) Classifying the soil as “Possible FILL” or “Probable FILL” is preferable.

The process of obtaining soil samples by means of the dynamic push technique disturbs the soil structure. Hence, the soil samples are disturbed samples and any tests affected by soil structure, such as instability index and strength, need to be treated with caution.

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Drillers are sub-professionals, often with little or no technical background.

Problems that can occur:

In some instances when stretch has occurred, the driller will omit some of the ‘stretched’ soil because the entire core sample will not fit neatly into the core box.

Sampling incorrect location or site;

Incorrectly documenting: tree location, direction of site slopes and other site features.

As a result, it is recommended practice by the author and many practicing geotechnical engineers, that an engineer be on-site during drilling. Whilst there are cost penalties associated with this, which can be passed on to the client, there are a number of significant benefits:

1. The engineer can ensure that the correct site and locations are being sampled;

2. If there is an obstruction in relation to one or more of the boreholes, the engineer can decide where the new borehole location is to be, rather than the driller;

3. The engineer can identify depths at which each tube sample ended and, hence, account for stretch and missing core;

4. The engineer can observe and note important site features, such as: tree locations, site drainage, evidence of fill and subsidence, depth to groundwater, cracking in existing or adjacent structures

6. SWELL PRESSURES It is well recognised that swelling can be suppressed when the surcharge load is large enough (Chen 1975). The swell pressure is measured using a constant volume or swell pressure test and involves inundating a sample in an oedometer while preventing the sample from swelling (Nelson and Miller 1992). The swell pressure is reported as the maximum applied stress required to maintain constant volume. When volume change is suppressed swell pressures as high as 1,000 kPa have been recorded, which is equivalent to a 40 to 50 m high embankment (Holtz and Kovacs 1981). Chen (1975) performed thousands of swell tests on a variety of soils from the northern United States. He found that: The swelling pressure of a clay is independent of the surcharge pressure (Fig. 6.1), initial

moisture content (Fig. 6.2), final degree of saturation (Fig. 6.3) and thickness of the stratum; and

The swelling pressure increases as the initial dry density grows (Fig. 6.4). He defined the swell pressure of an undisturbed soil as the pressure required to maintain its volume constant at its in situ dry density.

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0%

1%

2%

3%

4%

5%

6%

7%

8%

9%

10%

0 100 200 300 400

Surcharge Pressure (kPa)

Volu

me

Cha

nge

0

100

200

300

400

500

600

700

800

0 100 200 300 400

Surcharge Pressure (kPa)Sw

ell P

ress

ure

(kPa

)

(a) (b) Figure 6.1 Relationship between surcharge pressure and (a) volume change, and

(b) swell pressure. (After Chen 1975).

0%

1%

2%

3%

4%

5%

6%

7%

8%

9%

10%

0 5 10 15 20 25

Moisture Content (%)

Volu

me

Cha

nge

0

100

200

300

400

500

600

700

800

0 5 10 15 20 25

Moisture Content (%)

Swel

l Pre

ssur

e (k

Pa)

(a) (b)

Figure 6.2 Relationship between initial moisture content and (a) volume change, and (b) swell pressure. (After Chen 1975).

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0%

1%

2%

3%

4%

5%

6%

7%

8%

9%

10%

50 60 70 80 90 100

Degree of Saturation (%)

Volu

me

Cha

nge

0

100

200

300

400

500

600

700

800

900

50 60 70 80 90 100

Degree of Saturation (%)Sw

ell P

ress

ure

(kPa

)

(a) (b)

Figure 6. 3 Relationship between degree of saturation and (a) volume change, and (b) swell pressure. (After Chen 1975).

0%

1%

2%

3%

4%

5%

6%

7%

8%

9%

10%

1.0 1.2 1.4 1.6 1.8 2.0

Dry Density (t/m 3)

Volu

me

Cha

nge

0

200

400

600

800

1000

1200

1400

1600

1800

1.0 1.2 1.4 1.6 1.8 2.0

Dry Density (t/m 3)

Swel

l Pre

ssur

e (k

Pa)

(a) (b)

Figure 6.4 Relationship between initial dry density and (a) volume change, and (b) swell pressure. (After Chen 1975).

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7. SOIL SUCTION Soil suction is commonly referred to as the free energy state of soil water, which can be measured in terms of the partial vapour pressure of the soil water. The thermodynamic relationship between soil suction and the partial pressure of the porewater vapour can be written as follows (Fredlund and Rahardjo 1993):

⎟⎟⎠

⎞⎜⎜⎝

⎛−=ψ

00

lnv

v

vw uu

wvRT (7.1)

where: ψ is the soil or total suction (kPa);

R is the universal (molar) gas constant [= 8.31432 J/(mol K)]; T is the absolute temperature [i.e. T = (273.16 + t°) (K)] t° is the temperature (°C); vw0 is the specific volume of water or the inverse of the density of water

[i.e. 1/ρw (m3/kg)]; ρw is the density of water (i.e. 998 kg/m3 at t° = 20°C); wv is the molecular mass of water vapour (i.e. 18.016 kg/kmol); vu is the partial pressure of porewater vapour (kPa); and

0vu is the saturation pressure of water vapour over a flat surface of pure water at the same temperature (kPa).

The term 0vv uu is called the relative humidity, RH (%), and, if we select the reference temperature to be 20°C, Equation (7.1) simplifies to: ( )0ln022,135 vv uu−=ψ (7.2) A relative humidity value less than 100% in a soil would indicate the presence of suction. In its simplest terms, soil suction is the soil’s affinity for pure water. As we saw in Geotechnical Engineering II, total soil suction is the sum of two components matric (or matrix) suction and solute suction. 7.1 Matric Suction Matric suction results from the capillary action of the porewater. The voids between the individual particles of the soil form long, narrow and convoluted channels through which water is drawn. Surface tension of the water is the mechanism by which water is drawn up these narrow channels in the same way that water rises up a capillary tube and water is drawn by plants, as shown in Figure 7.1. Fundamentally, surface tension results from the differences in forces of attraction between the molecules of the materials at the interface. 7.2 Solute Suction Solute suction occurs where saline porewater exhibits an osmotic effect thereby attracting pure water. Figure 7.2 shows a bath which is divided in two by a semi-permeable membrane, that is, one which only permits the transfer of water molecules. On one side of the membrane is pure water and

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Figure 7.1 Matric suction.

Pure Water Saline Water

Na+

Na+Cl-

Cl-Water Flow

Semi-permeableMembrane

Figure 7.2 Osmotic pressure. on the other is a solution of saline water. In order to obtain equilibrium, the saline water will try and dilute itself by drawing water molecules from the pure water side of the bath. This process is known as osmosis and the pressure required to stop the flow through the semi-permeable membrane is known as the osmotic pressure or the solute suction. The semi-permeable membrane in clays is the strongly held cations near the surface of the clay particle. 7.3 Total Suction The total suction is simply the addition of matric and solute suction. Typical values of total suction are 150 kPa (pF 3.2) in temperate areas, and 1.5 MPa (pF 4.2) in semi-arid areas. A capillary rise of 5 m represents a matric suction of 5 × γw = 49 kPa. Clearly, in semi-arid areas, such as South Australia, the solute suction is the dominant effect. Soil suction is usually expressed in terms of pF (picofarads), kPa or MPa, where:

capillary rise

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⎟⎠

⎞⎜⎝

⎛=

0981.0log= cm)in water ofheight (log 1010

wPpF (7.1)

where: Pw is the water pressure expressed in terms of kPa. Thus a pF of 1 corresponds to 10 cm of water or 0.981 kPa. A dry soil has a greater affinity for water, that is a higher suction, and conversely, a wet soil has a lower suction. Total soil suction varies between pF 0 (wet) and ≈ 6 (dry) and is usually between pF 2 and 5. 7.4 Soil-Water Characteristic The relationship between soil suction and moisture content (Ww /Ws) is known as the soil-water characteristic, or moisture characteristic. Typical examples are shown in Figure 7.3. It is important to note from these plots that the soil-water characteristic is non-linear and unique for each soil. Hence, it is generally difficult, if not impossible, to predict soil suction from moisture content.

(a) (b)

Figure 7.3 Examples of soil-water characteristic curves from (a) Peter (1979a) and (b) Fredlund and Rahardjo (1993).

7.5 Measurement of Suction Total soil suction can be measured in the laboratory, and occasionally in the field, be the use of a number of different methods. These include the: psychrometer (thermocouple, transistor, or thermistor); vacuum desiccator; suction plate; filter paper; and the pressure membrane. These are described below.

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7.5.1 Psychrometer The thermocouple psychrometer is the most widely used technique to measure soil suction and uses a thermocouple to measure relative humidity, as shown in Equation (7.1). The most common thermocouple psychrometer is the Peltier type which uses the Peltier effect. In 1834, Peltier observed that when an electric current was passed through a circuit consisting of two dissimilar metals, as shown in Figure 7.4, one junction was cooler than the other. When current is passed in the opposite direction, the temperature gradient reverses.

Figure 7.4 Electric circuit to illustrate Peltier effect (Fredlund and Rahardjo 1993).

Peltier effect thermocouple psychrometers, such as the type manufactured by Wescor (Figure 7.5), can utilise either the wet bulb (psychometric) method, or the dew point (hygrometric) method. The wet bulb method involves measuring soil suction (water potential) by determining the wet bulb depression temperature. The thermocouple is cooled below the dew point by means of the Peltier effect, where micro-droplets of water condense on the junction surface. The water is then allowed to evaporate, causing the temperature of the junction to be depressed below the ambient temperature, due to evaporative cooling. The wet bulb temperature depression persists until all the water has been evaporated; then the thermocouple returns to the ambient temperature.

Figure 7.5 Wescor thermocouple psychrometer.

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The cooling coefficient , πv, of a thermocouple transducer is defined as the electromagnetic force, in microvolts, produced by the temperature differential that results from the passage of a specified nominally optimum cooling current through the junction. Hence, a microvoltmeter is read and the relative humidity inferred from a calibration curve. The dew point method involves measuring soil suction by determining the dew point depression temperature. The thermocouple is cooled below the dew point as in the wet bulb method, but the temperature of the thermocouple is then controlled by the heat of condensation from the water condensing on its surface. This causes the thermocouple temperature to converge to the dew point, where it remains with a static amount of water. The dew point measurement is therefore continuous in nature, rather than transitory, as with the wet bulb depression measurement. This allows greater measurement precision. The dew point method is specified by the Australian Standard AS 1289.2.2.1–1998 (Standards Australia 1998) and is suitable for measuring soil suctions between approximately pF 3.2 and 5. The procedure involves obtaining and intact sample of approximately 7 mm in diameter and 3 to 4 mm in thickness. The sample is then placed in a standard sample cup, 9.5 mm diameter × 4.5 mm deep (Figure 7.6), which is subsequently placed in the sample chamber, and the relative humidity read in terms of microvolts.

Figure 7.6 Sample chambers for Wescor thermocouple psychrometer. Decagon manufacture a dew point hygrometer (Figure 7.7) that uses the chilled-mirror dewpoint technique to measure the water potential of a sample. In this type of instrument, the sample is equilibrated with the headspace of a sealed chamber that contains a mirror and a means of detecting condensation on the mirror. The manufacturer’s claim that, under normal operating conditions, the apparatus can measure suctions within 5 minutes and with an accuracy of ±0.1 MPa, from 0 to –10 MPa (pF 5), and ±1%, from –10 to –60 MPa (pF 5.8). In situ thermocouple probes are also available (Figure 7.8), but have yet to be widely adopted by the geotechnical engineering profession. Becoming increasingly popular is transistor psychrometer (Figure 7.9) which uses two transistors which act as wet and dry bulb thermometers to measure the relative humidity of the air space in equilibrium with the soil sample. The temperature depression of the wet transistor, which holds a standard, manually-applied, water drop (Figure 7.10), is measured and amplified within the probe. As with the thermocouple psychrometer, the transistor psychrometer uses the relative humidity to determine the soil suction. The range of the transistor psychrometer is pF 3.0 to 5.0.

28

Figure 7.7 Decagon thermocouple psychrometer.

Figure 7.8 In situ thermocouple psychrometer probes.

Figure 7.9 Transistor psychrometer.

Wet transistor Dry transistor

29

(a)

(b)

Figure 7.10 Transistor psychrometer probe (a) diagram, (b) application of water drop. 7.5.2 Vacuum desiccator The vacuum desiccator makes use of a desiccant, a liquid of known suction, usually sulphuric acid or salt water (Figure 7.11). A vacuum is applied to the container so that water is able to diffuse more rapidly. The weight of the soil sample is measured periodically until it has stabilised. Soil suction equilibrium will be reached in approximately 1 to 2 weeks with an applied vacuum. By measuring the initial and moisture contents of the soil sample, it is possible to infer the initial soil suction. This does, however, rely on the assumption of a linear soil-water characteristic curve.

(a) (b)

Figure 7.11 Vacuum desiccator: (a) equipment, (b) diagram (Peter 1979b).

Soil

Printed circuit board

Water drop

30

7.5.3 Suction plate In the suction plate (Figure 7.12) a pressure difference is established across a membrane or porous plate by the application of direct suction to the underside while maintaining atmospheric pressure on the upper surface (Peter 1979b). The suction induced in the plate is transmitted to the soil sample placed upon it, and equilibrium is attained following any necessary flow of water.

(a) (b)

Figure 7.12 Suction plate method: (a) equipment (Sheard and Bowman 1996), (b) diagram (Peter 1979b).

7.5.4 Pressure membrane In the pressure membrane (Figure 7.13) the same principle is adopted as in the suction plate, with the difference being that the ambient air pressure is raised (Peter 1979b). For convenience of operation, a zero suction (or atmospheric pressure) is applied beneath the membrane, and the magnitude of the suction at the upper surface of the membrane is therefore directly determined by the air pressure in the sample chamber.

(a) (b)

Figure 7.13 Pressure membrane method: (a) equipment, (b) cell.

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7.5.5 Filter paper method In this technique the filter paper is used as a sensor and is based on the assumption that a filter paper will come into moisture flow equilibrium with the suction of the soil (Fredlund and Rahardjo 1993). Equilibrium can be attained with exchange of moisture in liquid or vapour form. Once equilibrium has been achieved, the moisture content of the filter paper is measured, which is compared with a calibration curve between suction and the moisture content of the filter paper.

(a) (b)

Figure 7.14 Filter paper method: (a) equipment, (b) sample setup. Other methods, such as tensiometers and gypsum blocks, are not presented here but are treated elsewhere (e.g. Fredlund and Rahardjo 1993). 7.6 Soil Suction Profiles Figure 7.15 shows typical soil suction profiles; that is, soil suction plotted as a function of depth. Figure 7.15(a) shows the typical shape of the soil suction profile beneath a well ventilated floor, such as a timber floor suspended on strip footings. Remembering that high suction suggests dry soil, as one would expect, the profile implies soil desiccation (drying out) at the ground surface asymptoting to moist soil at depth. Figure 7.15(b) shows the opposite trend beneath a well watered garden – wet at the surface, returning to moist soil at depth. Figure 7.15(c) shows soil suction profiles adjacent to and away from a tree. As shown, the tree’s roots dry out the soil at their depth. Figure 7.15(d) shows the suction profile, plotted against time, beneath an impervious barrier placed on the ground surface. At time, t = 0, the ground surface is drier than at depth. However, as time continues, the soil wets up as a result of the cessation of evaporation. Figure 7.15(e) shows how the climate influences soil suction. It also demonstrates that below a certain depth, approximately 4 m in Adelaide (Figure 7.16), soil suction is unaffected by climate.

32

Figure 7.15 Typical soil suction profiles.

(a) Suction under well ventilated floor (b) Suction under well watered garden

(c) Suction adjacent to and away from trees (d) Suction beneath an impervious

membrane as a function of time

(e) Suction with respect to the seasons in a semi-arid climate

33

Figure 7.16 Typical soil suction profiles (Mitchell and Avalle 1984). 8. MEASURES OF REACTIVITY It has been experimentally observed that the shrink, or swell, displacement is approximately linearly dependent on the change in total soil suction. The instability index of a soil, Ipt, is a measure of this relationship and is a measure of the degree of expansiveness of a soil. We define the instability index as:

u

I pt Δ

ε= (8.1)

where: ε vertical strain of the soil (= Δl/l, where l is the original height of the soil, and

Δl is the change in its height.); and Δu change in total suction. It is also experimentally observed that the instability index of a soil is related to the plasticity of the soil, as shown in Table 8.1. As is evident, as the plasticity – indicated by the plasticity index, PI – increases, so too does the reactivity of the soil. Figure 8.1 shows the relationship between the instability index and the plasticity index of the Adelaide soil profiles discussed previously in §4.

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Table 8.1 Approximate instability indices, Ipt, with respect to plasticity index, PI.

Description of Plasticity

Plasticity Index (%)

Instability Index (%)

Trace < 2 < 0.5 Very Low 2 to 5 Approx. 0.5

Low 5 to 10 0.5 to 1 Low to Medium 10 to 20 1 to 1.5

Medium 20 to 25 1.5 to 2 Medium to High 25 to 30 Approx. 2

High 30 to 45 2 to 3.5 Very High 45 to 60 3.5 to 5

Extremely High > 60 > 5

Figure 8.1 Relationship between Ipt and PI for Adelaide soils (Mitchell and Avalle 1984). 8.1 Measurement of Instability Index AS 2870 recommends three direct soil tests for determining the instability index of a soil: the core shrinkage test; loaded shrinkage test; and shrink-swell test. 8.1.1 Core shrinkage test The core shrinkage test (AS 1289.7.1.3–1998), as shown in Figure 9.1(a), consists of trimming the ends of a small, undisturbed core of soil, usually 38 to 50 mm in diameter. A drawing pin is located at the centre of each end to enable measurement of changes in core length. The initial suction of the

35

sample is determined from the sample trimmings using one of the methods described in §7.5, above; usually the psychrometer. The core sample is allowed to air-dry for at least two days and is the oven-dried to determine the final moisture content. It is assumed that the air-dry suction is equal to pF 6.8, as shown in Figure 9.1(b) (Mitchell and Avalle 1984). Mass and length of the sample are frequently monitored over the first two days to enable construction of strain-moisture content curves. The core shrinkage index, Ics, is then calculated using the initially linear part of the drying curve [Figure 9.2(a)], and is given by Equation (8.2) and Figure 9.2(b).

uw

wuIcs Δ

Δ

Δ

ε=

Δ

ε= . (8.2)

where: ε corresponding vertical strain; Δu change in total suction (pF), and; Δw change in moisture content (%).

(a) (b)

Figure 9.1 Core shrinkage test: (a) set up; (b) assumption of pF 6.8 at zero moisture content.

(a) (b)

Figure 9.2 Core shrinkage test results: moisture content vs. (a) strain; (b) suction.

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8.1.2 Loaded shrinkage test This test uses a perforated shrinkage cell which accommodates a small sample, loaded by adjustable springs, as shown in Figure 9.2. A small load of 25 kPa is normally applied to the soil for house foundation investigations. The initial moisture content and suction of the sample need to be determined. The end moisture condition is achieved by confining the cell in a vacuum desiccator over super-saturated, copper sulfate solution. The humidity within the desiccator slowly establishes an equilibrium sample suction approaching pF 4.5, which can be verified after mass equilibrium is reached. The loaded shrinkage index, Ils, can then be determined. This test usually takes many weeks to complete.

Figure 9.2 Loaded shrinkage test. 8.1.3 Shrink-swell test Cameron (1989) stated that, generally, the assessment of reactivity by loaded swell tests alone has been discouraged because of the difficulty in interpreting results of any worth to actual site conditions. Use of distilled water has an appreciable effect on solute suction and hence volume change in saline soils. Furthermore, vertical swelling strains are exaggerated by the imposition of complete lateral restraint using a rigid sample ring. In the shrink-swell test (AS 1289.7.1.1–2003), as shown in Figure 9.3, the disadvantages of swell testing are reduced, firstly by applying an empirical correction factor to the swelling strain and, secondly by combining it with a simplified core shrinkage test on a companion sample at the same field moisture condition. The total swelling strain is reduced by a factor of two to account for the effect of the rigid sample ring. The shrink-swell index, Iss, is then given by:

( )⎥⎦

⎤⎢⎣

⎡ ε+ε=

8.12 shsw

ssI (8.2)

37

where εsw is the swelling strain and εsh is the shrinking strain to the oven-dry condition. The denominator of 1.8 represents the likely suction range over which volume changes are almost linearly proportional to the suction changes.

Figure 9.3 Shrink-swell test. The shrink-swell test has two considerable advantages over the other tests. Firstly it does not require any suction testing and secondly, it can be used to assess reactivity for soil at any given initial moisture condition. This test usually takes around one week to complete. Each of the three tests described, give a different value for the Ipt, with the core shrinkage test being the least reliable (Cameron 1989). A simple and universal instability index test has yet to be found. However, in the majority of cases the visual-tactile method is used, which involves a visual and manual inspection of the soil core, in combination with Table 8.1. The paper by Jaksa et al. (1997), included at the end of these notes, details an assessment of the variability of the visual-tactile method. 9. HEAVE It has been proposed, that a soil layer of thickness, t, and instability index, Ipt, which is subjected to a change in total suction, Δu, will undergo a heave, y, that is, a displacement (shrink or swell), equal to that given by Equation 3.1. tuIy pt ..Δ= (9.1) For a layered soil profile, we sum the individual heaves for each layer, as shown in Equation 3.2.

( )∑=

Δ=N

iiipts tuIy

i1

.. (9.2)

where: ys total vertical surface heave, that is, the vertical heave theoretically observed at

the ground surface; N total number of soil layers influenced by the change in soil suction; i the layer number.

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Example 1. Given the soil profile and total suction change profile as shown in Figure 9.1, calculate the total vertical surface heave.

Figure 9.1 Example 1. Solution:

39

9.1 Classification to AS 2870–2011 The Australian Standard “Residential Slabs and Footings”, AS 2870–2011 (Standard Australia 2011), stipulates the following site classification system (Table 9.1), based on the characteristic surface movement, ys. The characteristic surface movement is defined as “the movement of the surface of a reactive site caused by moisture changes from characteristic dry to characteristic wet condition in the absence of a building and without consideration of load effects.” The characteristic surface movement is calculated using Equation (9.2), Table 9.2 and Figure 9.2. Table 9.1 Classification by characteristic surface movement (AS 2870–2011).

Table 9.2 Classification by site reactivity (AS 2870–2011).

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Figure 9.2 Effect of bedrock or watertable on design suction change profiles (AS 2870–2011). In addition, AS 2870 stipulates that “for Classes M, H1, H2, and E further division based on depth of the expected movement is required. For deep-seated movements, characteristic of dry climates and corresponding to a design depth of suction change, Hs, equal to or greater than 3 m, the classification shall be M-D, H1-D, H2-D or E-D as appropriate. Recommended soil suction change profiles for several locations in Australia are given in Table 9.3. Table 9.3 Recommended soil suction change profiles for certain locations (AS 2870–2011).

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10. REFERENCES Cameron, D. A. (1989). Test for Reactivity and Prediction of Ground Movement. Civil Engineering Transactions,

I.E.Aust., 121–132. Chen, F. H. (1975). Foundations on Expansive Soils, Elsevier, Amsterdam. Fredlund, D. G. and Rahardjo, H. (1993). Soil Mechanics for Unsaturated Soils, Wiley, New York. Holtz, R. D. and Kovacs, W. D. (1981). An Introduction to Geotechnical Engineering, Prentice Hall, New Jersey. Lambe, T. W. and Whitman, R. V. (1979). Soil Mechanics, S.I. Version, Wiley, New York. Mitchell, J. K. (1976). Fundamentals of Soil Behaviour, John Wiley and Sons, New York. Mitchell, P. W. and Avalle, D. L. (1984). A Technique to Predict Expansive Soil Movements. Proc. 4th Int. Conf. on

Expansive Soils, Adelaide, 124–130. Nelson, J. D. and Miller, D. J. (1992). Expansive Soils: Problems and Practice in Foundation and Pavement

Engineering, John Wiley and Sons, New York. Peter, P. (1979a). Soil Moisture Suction. In Footings and Foundations for Small Buildings in Arid Climates, I.E.Aust,

46–62. Peter, P. (1979b). Laboratory Methods. In Footings and Foundations for Small Buildings in Arid Climates, I.E.Aust,

90–96. Selby, J. and Lindsay, J. M. (1982). Engineering Geology of the Adelaide City Area. S.A. Dept. Mines and Energy

Bulletin 51, Adelaide. Sheard, M. J. and Bowman, G. M. (1996) Soils, Stratigraphy and Engineering Geology of Near Surface Materials of

the Adelaide Plains. Report Book 94/9, DME 565/79, 3 Vols. Standards Australia (1998). AS 1289.2.2.1–1998 Standards Australia (2011). Residential Slabs and Footings – Construction, AS 2870. Stapledon, D. H. (1970). Changes and Structural Defects Developed in Some South Australian Clays, and their

Engineering Consequences. Symp. on Soils and Earth Structures in Arid Climates, Inst. Eng., Aust. and Aust. Geomech. Soc., Adelaide, 62–71.

Steinberg, M. (1998). Geomembranes and the Control of Expansive Soils in Construction, McGraw-Hill, New York. Taylor, J. K., Thomson, B. P. and Shepherd, R. G. (1974). The Soils and Geology of the Adelaide Area. Bulletin

46, Dept. of Mines.

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UNCERTAINTIES ASSOCIATED WITH THE VISUAL-TACTILE METHOD FOR QUANTIFYING THE REACTIVITY OF EXPANSIVE SOILS

M. B. Jaksa Lecturer, Department of Civil and Environmental Engineering, The University of Adelaide

R. L. Cavagnaro Geotechnical Engineer, Department of Transport, South Australia

D. A. Cameron Senior Lecturer, School of Engineering, University of South Australia

Abstract: Quantifying the instability index, Ipt , of the various soil horizons within the subsurface profile is central to the design of residential foundations built on expansive soils. The visual-tactile method for estimating Ipt (Standards Australia, 1981; Mitchell, 1989; Standards Australia, 1996) has been widely adopted throughout the geotechnical engineering community. However, by its very nature, the method is highly classifier-dependent. A joint study between The University of Adelaide, the Footings Group of the S.A. Division of the Institution of Engineers, Australia and the University of South Australia has been conducted in an attempt to quantify the uncertainties associated with the visual-tactile method. Continuous boreholes were drilled at three sites in metropolitan Adelaide and the core samples were identified by 14 different engineers practicing in the design of residential footings. A series of Atterberg limit and shrink-swell tests were performed in order to classify the soil layers and to benchmark the instability indices. It has been found that the estimated values of Ipt , and the resulting estimates of design surface movement, ys , vary markedly and are highly classifier-dependent.

1. INTRODUCTION The design of residential foundations built on expansive soils is based on an estimate of the design surface movement, ys , which is given by the following expression (Aitchison, 1973):

(1)

where: Ipt is the instability index of the soil, which is defined as the percent vertical strain per unit change in suction;

Δu is the change in suction, in pF units, at a depth z below the ground surface; Δh is the thickness of the soil layer under consideration; and Hs is the depth of the design suction change.

Central to the calculation of ys are estimates of the instability indices, Ipt , of the underlying soil profile. The Australian Standard for the design and construction of residential slabs and footings, AS 2870-1996 (Standards Australia, 1996), specifies three methods for the estimation of Ipt : 1. Laboratory tests. Three such tests are suggested: the shrink-swell test, AS 1289.7.1.1-1992; the loaded shrinkage

test, AS 1289.7.1.2-1992; and the core shrinkage test, AS 1289.7.1.3-1992 (Standards Australia, 1992).

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2. Correlations between the shrinkage index, Ips , and other clay index tests; and 3. Visual-tactile identification of the soil by an engineer or engineering geologist having appropriate expertise and

local experience. This paper focuses on the third technique, namely visual-tactile identification. This technique, also referred to as the visual-manual method (Mitchell, 1989), involves a visual inspection of the soil and manually moulding and kneading the soil in order to estimate its Plasticity Index, PI. Mitchell (1979) presented an approximate relationship between PI and Ipt which is often used in the visual-tactile method to estimate Ipt . The visual inspection assists in the estimation of Ipt by identifying any structure within the soil sample, giving an appreciation of the pedological soil type (Taylor et al., 1974), as well as indicating the proportion of non-reactive inclusions, such as sand and gravel. The implementation of the visual-tactile method of Ipt estimation, however, appears to vary between practitioners, which is likely to contribute to the variability of the resulting estimates.

2. STUDY In an effort to quantify the uncertainty associated with the visual-tactile method, the authors coordinated a study which involved obtaining a number of continuous soil samples, distributing these samples to several geotechnical engineering consultants for classification and Ipt estimation, and comparing these with laboratory measurements of Ipt . Three sites within metropolitan Adelaide were identified for investigation, as detailed in Table 2.1. These sites were chosen to provide a variety of soil types and classes of site reactivity. Site No.

Suburb Within Adelaide

Site Description

Pedological Soil Type (Taylor, 1974)

Expected Site Classification

1 Eastwood Vacant lot, level Red-brown earth (RB3) H 2 Woodcroft Vacant lot, gentle slope Black earth (BE) E2 3 Blackwood Vacant lot, moderate slope Podsolic Soil (P2) M

Table 2.1 Site characteristics. At each of the three sites, two, 40 mm diameter boreholes were drilled, adjacent to one another and to a depth of 3 metres, by means of the dynamic push technique. As a result, two, practically identical continuous core samples were obtained for each site. This enabled the samples to be identified and classified by the geotechnical engineering consultants, within a relatively short time frame, without the soil drying-out to any great extent, and thereby affecting the results. In all, 14 geotechnical engineers from the following consulting practices and educational institutions participated in the study: Acer Wargon Chapman; B. K. Andrews & Partners; Bastick & Partners; Coffey Partners International; Department of Civil & Environmental Engineering, The University of Adelaide; M. R. Herriot & Associates; Trevor John & Associates; Koukourou & Partners; Rust PPK; John Sandland & Associates; B. C. Tonkin & Associates; and TMK & Associates. Each geotechnical engineer was requested to submit a borelog for each of the three sites which included the following data: an estimate of the instability index, Ipt , for each soil layer; the depths to the interfaces of each horizon; and a description and Unified Soil Classification for each soil layer. In addition, at the Eastwood and Woodcroft sites, a number of 90 mm diameter, undisturbed samples were obtained to enable laboratory measurements of Ipt to be determined. Due to problems associated with the drilling equipment, no undisturbed samples were obtained from the Blackwood site.

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3. LABORATORY RESULTS In order to be able to compare the geotechnical engineers’ estimates of Ipt with values measured in the laboratory, a number of shrink-swell tests were performed on undisturbed, 90 mm diameter, samples. The tests were performed in accordance with AS 1289.7.1.1-1992 (Standards Australia, 1992). While the three laboratory tests specified by AS 1289, mentioned previously, each have varying degrees of reliability, Cameron (1989) suggested that the shrink-swell approach appears to be the most successful method. It was for this reason that the shrink-swell test was adopted. Measured values of the shrink-swell index, Iss, for the Eastwood and Woodcroft sites are given in Table 3.1. In addition to the shrink-swell tests Atterberg limits tests were carried out to determine the liquid limit, wL , plastic limit, wP , and the Plasticity Index, PI. These results are also given in Table 3.1. Figure 3.1 shows the measured relationship between Iss and PI. A line of best fit, obtained by ordinary least squares regression of a line passing through the origin, is shown superimposed on this graph. While only four test results were obtained, there appears to be relatively good correlation between the two variables, as is indicated by the coefficient of determination, r2, being close to unity. Table 3.1 Laboratory test results.

Site Description Depth (m)

wL (%) wP (%) PI (%) Iss (%)

Eastwood Clay, CH, red-brown 0.5-0.8 78.0 27.8 50.2 5.4 Clay, CH, red-brown, 30% calc. material 0.8-1.3 - - - 2.7 Clay, CH, red-brown, 5% calc. material 1.8-2.3 51.5 27.5 24.0 2.1 Clay, CH, red-brown, 5% calc. material 3.0-3.1 47.5 17.5 30.0 -

Woodcroft Clay, CH, black, 10% calc. material 0.2-0.7 45.8 8.2 37.6 2.6 Clay, CH, grey, 30% calc. material 1.1-1.2 67.2 30.4 36.8 - Clay, CH, grey-green 2.0-2.5 75.0 23.4 51.6 5.6

Blackwood Clay, CH, red-brown 1.8-2.0 54.2 25.4 28.8 - Completely weathered siltstone 2.6-2.8 48.5 40.9 7.6 -

1

1

1

1

0

1

2

3

4

5

6

0 10 20 30 40 50 60

Plasticity Index, PI (%)

Line of Best Fit(r2 = 0.89)

Mitchell (1979)Relationship

Figure 3.1 Experimental relationship between Iss (Iptt) and PI.

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The approximate relationship proposed by Mitchell (1979), which is used by some practitioners in the visual-tactile method, is also shown in Figure 3.1. There seems to be relatively good agreement between the line of best fit and Mitchell’s relationship. It is important to note, however, that Mitchell and Avalle (1984) showed that, even within pedological soil groups, considerable scatter still exists between Ipt and PI.

4. SURVEY RESULTS The results of the visual-tactile method survey are summarised in Figures 4.1 to 4.3 for the Eastwood, Woodcroft and Blackwood sites, respectively. The complete survey submissions are given by Eden and Hill (1994). Superimposed on Figures 4.1 and 4.2 are the measured values of Iss , discussed in the previous section, minimum and maximum envelopes, as well as the mean of the estimates. It is evident from these figures that there is large variability in the estimates of Ipt as specified by the geotechnical engineers. It is interesting to note that 2 of the 5 laboratory measured values of Ipt fell outside of the envelope of the geotechnical engineers’ estimates, indicating a relatively low level of accuracy associated with the visual-tactile procedure.

J

J

J

3

2.5

2

1.5

1

0.5

0

0 1 2 3 4 5 6 7

Estimated Instability Index, Ipt (%)

Max.

Min.

Mean

Measured values

Figure 4.1 Variation of estimated Ipt with depth, Eastwood site. The design of residential footings is greatly influenced by the design surface movement, ys , which is evaluated using Equation (1). Since ys is dependent, not only on estimates of Ipt , but also on the thickness of the various layers and the adopted soil-suction profile, it is more appropriate to study the variability of the visual-tactile method by examining the variability in ys . It should be noted, however, that, since it is common practice to adopt a standard triangular soil-suction profile (Standards Australia, 1996), the variability in ys from the visual-tactile procedure, is dependent on the variability in Ipt and layer thickness, alone. The variability in ys , for each of the three sites, is shown in Figures 4.4 to 4.6. It is apparent from these histograms that there is large scatter in the derived estimates of design surface heave. This scatter is quantified by the statistics given in Table 4.1. In particular, the percent variation, of between 99% and 166%, indicates significant scatter associated with the derived values of ys from the visual-tactile method, as does the coefficient of variation of between 17.4% and 29.0%.

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J

J

3

2.5

2

1.5

1

0.5

0

0 1 2 3 4 5 6 7 8 9 10

Estimated Instability Index, Ipt (%)

Max.

Mean

Measured values

Min.

Figure 4.2 Variation of estimated Ipt with depth, Woodcroft site.

3

2.5

2

1.5

1

0.5

0

0 1 2 3 4 5 6

Estimated Instability Index, Ipt (%)

Max.

Mean

Min.

Figure 4.3 Variation of estimated Ipt with depth, Blackwood site.

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40 50 60 70 80 900

5

10

15

20

25

30

30Calculated Surface Heave, ys (mm)

M H E1Site Classifications

Figure 4.4 Histogram of estimated Ipt , Eastwood site.

80 90 100 110 120 130 140 150 160 1700

5

10

15

20

25

30

35

40

45

70Calculated Surface Heave, ys (mm)

E2Site Classifications

E1

Figure 4.5 Histogram of estimated Ipt , Woodcroft site.

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10 20 30 400

10

20

30

40

50

60

30Calculated Surface Heave, ys (mm)

MSSite Classifications

Figure 4.6 Histogram of estimated Ipt , Blackwood site.

Site Range, ysmin to

ysmax, (mm)

Variation (%)

Mean µµ , (mm)

Standard Deviation, σσ , (mm2)

Coefficient of Variation, CV, (%)

Eastwood 33 − 87.7 166 63.4 14.0 22.1 Woodcroft 84 − 167.3 99 116.7 20.4 17.4 Blackwood 13.9 − 34.7 150 22.1 6.4 29.0

Table 4.1 Statistics of derived values of design surface movement, ys , from the visual-tactile method survey.

(Note: and )

The variation and inaccuracy associated with the visual-tactile method is derived from a number of sources. Firstly, the method appears to vary among geotechnical engineering practitioners. This is likely to be due to the fact that the method is not stipulated in any of the literature. Secondly, it is an important requirement of the method that the classifiers’ estimates of Ipt be regularly calibrated against laboratory tests. In the vast majority of cases this is not done. Thirdly, some of the classifiers who participated in the study, have had little experience in geotechnical soil identification and classification, though they carry out this role for their respective companies from time-to-time. As a consequence, their estimates are likely to be less reliable, and more variable, than those from more experienced classifiers. Finally, the estimates of Ipt derived from the visual-tactile method often incorporate some level of scaling, dependent on the amount of risk that the classifier is prepared to accept. For example, if the classifier is relatively inexperienced, he/she is likely to be more cautious and, hence, attribute a higher value of Ipt than one who is more experienced and more confident with their estimates. Insufficient soil samples were obtained to enable the ‘true’ values of design surface heave, ys to be determined, at each of the three sites. As a consequence, it is not possible to compare the values of ys from the survey with those derived from laboratory measurements. Nevertheless, the survey provides a valuable insight into the accuracy, or otherwise, of the visual-tactile procedure.

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5. CONCLUSIONS This paper has presented the results of a study which has aimed to quantify the uncertainty associated with the visual-tactile method of instability index (Ipt ) estimation. It has been shown that this technique is highly variable, with values of design surface movement, ys , derived from the estimates of Ipt , varying by as much as 166%, and with a coefficient of variation of 29%. As a result, it is concluded that the visual-tactile method is unreliable and inaccurate, particularly when practitioners’ estimates are not regularly calibrated against laboratory tests. It is recommended that the method be further developed and standardised, or alternative techniques be adopted.

6. ACKNOWLEDGEMENTS The research detailed in this paper was undertaken in 1994 as part of a final year undergraduate project by C. M. Eden and C. J. Hill, under the supervision of M. B. Jaksa. The authors are grateful to Messrs. Eden and Hill for their contribution to this project. The generous support of the various geotechnical engineering organisations listed in Section 2 is gratefully acknowledged. In addition, the authors wish to thank the Footings Group of the S.A. Division of the Institution of Engineers, Australia, Mr. Tad Sawosko, Department of Civil & Environmental Engineering, The University of Adelaide, and the University of South Australia.

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