ANALYSIS OF INSITU TEST DERIVED SOIL PROPERTIES WITHTRADITIONAL AND FINITE ELEMENT METHODS
By
LANDY HARIVONY RAHELISON
A DISSERTATION PRESENTED TO THE GRADUATE SCHOOLOF THE UNIVERSITY OF FLORIDA IN PARTIAL FULFILLMENT
OF THE REQUIREMENTS FOR THE DEGREE OFDOCTOR OF PHILOSOPHY
UNIVERSITY OF FLORIDA
2002
To my Father, and my Mother
iii
ACKNOWLEDGMENTS
The author is most thankful to Dr. Frank Townsend for his guidance and support
throughout this research, having worked with Dr. Townsend during these three years has
brought the author very instructive experiences inside and outside the geotechnical area.
The author also owes his greatest gratitude to the committee members: Dr. Joseph
Tedesco for sacrificing his time and for his critical insights into the research, Dr. John
Davidson for his ideal academic instructions and for being a role model in and out of the
classroom, Dr. Paul Bullock for the invaluable assistance in the field testing and the
extremely instructive discussions in his office, and Dr. Steven Detweiler for his
contribution to this research by expressing different points of view.
The author is also indebted to the personnel of the Florida Department of
Transportation, especially the project coordinator Peter Lai, for their support and
technical assistance.
The author would like to express his gratitude to the personnel of Turner, Inc. at
the University of South Florida site, the personnel of Applied Foundations Testing, Inc.,
especially Mike Muchard and Don Robertson, and the personnel of Pile Equipment, Inc.
at the Green Cove Springs site.
The author would like to thank all those at the University of Florida, especially
Brian Anderson, Chris Kohlhof, Danny Brown, Bob Konz, Hubert Martin, and Tony
Murphy who assisted in many ways throughout this work.
iv
The author appreciates the unforgettable help from the Geotech graduate students
Jason Gowland, Joshua Logan, and Scott Jacobs.
The Geotech Fall ’99 crew, especially Rodrigo Herrera, Victor Alvarez, Michael
Kim, and Thai Nguyen will always be remembered for the exchange of knowledge and
the valuable discussions on daily events.
The author also thanks the younger and exuberant Geotech graduate students,
especially Zhihong Hu and Minh Le for improving his computer skills during the writing
of this dissertation.
The author is grateful to professors M. Sezaki, H. Yokota, and F. Imai of the
Department of Civil and Environmental Engineering, Miyazaki University, Japan, to
professor N. Cristescu of the University of Florida, and to the staff of the Agency of
Daiho Corporation in Madagascar.
Last but certainly not least, the author would like to present his most heartfelt
acknowledgements to his father for education and principles, to his mother for caring and
understanding, to his brothers and sisters for sound advice and for cheering him up all the
time, and to his long-time best friend Noriko Goto for encouragement, support and
patience.
v
TABLE OF CONTENTSpage
ACKNOWLEDGMENTS ................................................................................................. iii
ABSTRACT..................................................................................................................... viii
CHAPTER
1 INTRODUCTION ...........................................................................................................1
Background..................................................................................................................... 1Present Status .................................................................................................................. 3Objectives of Study......................................................................................................... 6Outline of Thesis............................................................................................................. 8
2 TRADITIONAL METHODS AND FEM CODE PLAXIS.............................................10
Sheet Pile Wall Computations with CWALSHT.......................................................... 10Stress Paths in Sheet Pile Walls.................................................................................... 16Settlement Calculations with CSANDSET................................................................... 17Settlement Calculations Using Insitu Tests .................................................................. 22Stress Paths in Shallow Footings .................................................................................. 24Statnamic Test and Analysis with SAW R4 ................................................................. 25Finite Element Code: PLAXIS........................................................................................ 30
3 LITERATURE REVIEW FOR INSITU TESTING......................................................56
General.......................................................................................................................... 56Standard Penetration Test: SPT .................................................................................... 57Cone Penetration Test: CPT.......................................................................................... 70Flat Dilatometer Test: DMT ......................................................................................... 83Pressuremeter Test: PMT.............................................................................................. 96Summary on Insitu Testing......................................................................................... 111
4 SHEET PILE WALL AT MOFFITT CANCER CENTER.........................................113
Introduction................................................................................................................. 113Site Description and Insitu Testing............................................................................. 115Sheet-Pile Wall Test Section ...................................................................................... 122Soil-Structure Profile for CWALSHT and FEM Modeling........................................ 125Slope Inclinometer Data ............................................................................................. 126
vi
Modeling with CWALSHT and FEM Code PLAXIS................................................... 131Discussion on Finite Element Modeling..................................................................... 143Conclusions................................................................................................................. 149
5 STUDY OF OTHER SHEET PILE WALLS ..............................................................152
General........................................................................................................................ 152Hochstetten Sheet Pile Wall........................................................................................ 152Hatfield Sheet Pile Wall.............................................................................................. 174Rotterdam Strutted Sheet Pile Walls in Very Soft Clay ............................................. 189Conclusion for Sheet Pile Walls ................................................................................. 209
6 CIRCULAR FOOTING AT GREEN COVE SPRINGS.............................................211
Introduction................................................................................................................. 211Site Description and Insitu Testing............................................................................. 213Static Load Testing ..................................................................................................... 220Modeling with CSANDSET and FEM code PLAXIS .................................................. 231Estimate of Bearing Capacity ..................................................................................... 242Settlement Results....................................................................................................... 245Conclusions................................................................................................................. 253
7 STUDY OF OTHER SHALLOW FOOTING CASES ...............................................256
General........................................................................................................................ 256Two Square Concrete Footings in Texas A&M University........................................ 256Power Plant Mat Foundation in Utah ......................................................................... 281Conclusion for Shallow Footings ............................................................................... 290
8 STATNAMIC LOAD TESTS ON SHALLOW FOUNDATIONS.............................292
General........................................................................................................................ 292Statnamic Load Testing .............................................................................................. 292Statnamic Test on Green Cove Springs Footing......................................................... 299Orlando Steel Plate Footing ........................................................................................ 313Conclusion for Statnamic Load Tests ......................................................................... 322
9 CONCLUSIONS AND RECOMMENDATIONS ......................................................324
Sheet Pile Wall Problems............................................................................................ 325Shallow Footing Problems.......................................................................................... 325Statnamic Problems .................................................................................................... 326Recommendations and Future Work .......................................................................... 327
vii
APPENDIX
A BASIC EQUATIONS OF THE DIFFERENT METHODS .......................................328
Traditional Method with CWALSHT......................................................................... 328Traditional Methods with CSANDSET...................................................................... 330Settlement with Insitu Tests........................................................................................ 332Statnamic Test: Unloading Point Method................................................................... 334
B SHEET PILE AT MOFFITT CANCER CENTER.....................................................336
Soil Properties from CPT, Moffitt Cancer Center ...................................................... 336Data Reduction in Slope Inclinometer Test ................................................................ 341
C SHEET PILE WALLS IN KARLSRUHE, HATFIELD AND ROTTERDAM.........344
D GREEN COVE SPRINGS SHALLOW FOOTING...................................................353
E TEXAS A&M FOOTINGS AND UTAH MAT FOUNDATION..............................359
LIST OF REFERENCES.................................................................................................371
BIOGRAPHICAL SKETCH ...........................................................................................381
viii
Abstract of Dissertation Presented to the Graduate Schoolof the University of Florida in Partial Fulfillment of theRequirements for the Degree of Doctor of Philosophy
ANALYSIS OF INSITU TEST DERIVED SOIL PROPERTIES WITHTRADITIONAL AND FINITE ELEMENT METHODS
By
Landy Harivony Rahelison
December 2002
Chair: Frank C. TownsendDepartment: Civil and Coastal Engineering
The strength and the stiffness properties of soils from four different insitu tests are
examined by studying three types of geotechnical engineering problems: unloading in
sheet pile walls; loading in shallow footings; and dynamic loading in Statnamic load test.
Two types of calculations were used: the traditional methods, and the FEM with PLAXIS.
The results of the analyses greatly depend on the choice of the insitu test and the
correlations selected for the soil properties. The analyses were based on the comparison
of the measured responses with those predicted.
For cantilevered, strutted, or anchored sheet pile walls, embedded in sands and
clays, the SPT test, PMT, and the CPT tests are the preferred insitu tests for design
predictions. The equivalent blow count CPT (N) also constitutes a very useful method in
estimating the soil properties. The traditional methods did not prove useful, as poor
predictions were obtained, and thus are not recommended. The FEM with PLAXIS has
better capability and versatility to analyze sheet pile wall problems. Apart from the
ix
efficacy of the Hardening Soil model, a good finding was the use of the unload-reload
modulus in the Mohr-Coulomb model. The Hardening Soil model is also appropriate to
simulate the behavior of very soft clays.
For shallow footings, knowledge of the stress history of sand layers proves highly
critical in obtaining good settlement predictions; the insitu tests DMT and CPT are
mandatory for estimating the OCR values and/or the K0 values. The SPT, CPT (N) based
traditional methods: D’Appolonia (1970), Schultz & Sherif (1973), and DMT method
provide the best predictions. The Mohr-Coulomb model using the CPT data or DMT
data is preferable for FEM predictions.
For the Statnamic load test, the Unloading Point Method was shown as a valid
alternative to the actual static load test. The method produced the closest load-settlement
curve to the actual static test. The CPT (N) data are useful in estimating the dynamic
properties of the soil. The FEM code PLAXIS successfully simulated the Statnamic test.
1
CHAPTER 1INTRODUCTION
Background
Geotechnical engineers deal with soil, the most challenging of civil engineering
materials. All civil engineering structures are built in, or on this complex material. Soil
exhibits a wide variety of physical, chemical, and mechanical properties. It is porous and
multiphased, formed by solid particles, gases, and water with irregular patterns. It is
inherently heterogeneous and discontinous, each individual particle being different from
the next. However, in macrosopic scale, a soil mass is usually characterized as a
homogeneous and solid continuum, although no real soil can be considered as a
continuous medium. A closer microspic look reveals various types of water surrounding
the solid particles: adsorbed water, double-layer water and free water (Das, 1998). The
presence of water in the pores greatly affects the compressibility of a soil mass (Terzaghi,
1943). Soil is formed through the combination of various geological, environmental, and
chemical processes (Kulhawy and Mayne, 1990). Based upon their behavior, soils can be
divided into two major groups: cohesive (clay) and cohesionless (sand). Unlike other
civil engineering materials such as concrete and steel, soil strength and stiffness
properties are not known in advance and must be evaluated only after conducting tests.
These properties vary widely from one site to another. To simplify testing and analysis,
soil is assumed primarily as isotropic. Generally soil behaves anisotropically. That is
properties such as elastic modulus E or hydraulic conductivity k, … are usually different
in the vertical direction than in the horizontal direction.
2
Soil is characterized by nonlinear stress-dependency of the stiffness properties. In
solving geotechnical problems, linear and nonlinear elastic (e.g., Duncan Chang)
relationships between stress and strain have been used (Lambe & Whitman, 1969) and
found useful especially for small deformation problems and dynamic problems.
Nevertheless, in many other cases, they do not duplicate closely the actual behavior of the
soil. Soil is plastic: clay, physically and mechanically, is a good example of a plastic
material (Lubliner, 1990). Simulation of the plastic behavior of the soil cannot be
achieved by pure mathematical formulation. In order to cover the gap between the actual
behavior and the various theoretical formulations, researchers and engineers have
attempted to introduce correction factors in their mathematical formulations to obtain
better results.
The two main problems a geotechnical engineer faces are to avoid stability failure
and to control deformations, with a certain safety factor. Examples of traditional
equations to analyze the stability of a shallow footing and a retaining structure are
discussed below.
γγγ NBNDNcq qfcULT ××+××+×=21 (1.1)
−××=2
45tan21 22 φ
γ HPa (1.2)
Equation (1.1) calculates the Bearing Capacity of a continuous footing of width B
(Terzaghi, 1967), and is based on the equilibrium criteria of soil surrounding the footing.
Similarly, Eq. (1.2) is computes the force from the active earth pressure behind a
retaining wall based on Rankine’s theory (Lambe & Whitman, 1969). Similar equations
3
also exist in the case of load-deformation analysis. They are either empirical equations or
are based on Elastic theory.
MBq ×∆
××= 21 µµρ (1.3)
2
122
+×
×=
BB
NPCDρ (1.4)
Equation (1.3), by D’Appolonia et al. (1968) is used to estimate the settlement ρ of a
shallow footing of width B under the load ∆q on sand whose compressibility is
characterized by the constrained modulus M. Correction factors µ1 and µ2 were needed
for better agreement with the actual cases. Equation (1.4) is used to compute the
settlement of a footing on sand using the Standard Penetration Test blow count N,
(Meyerhof, 1965) and uses the embedment correction factor CD.
Equations (1.1) and (1.2) are based on plastic equilibrium theory. Equation (1.3)
is based on the theory of elasticity. Equation (1.4) is an empirical equation. These
equations, still widely used in geotechnical engineering, belong to the group of the
traditional methods designated in this study. The traditional methods are readily
available because of the advent of computer programs.
Present Status
Geotechnical engineers have developed various computer programs to solve the
types of problem mentioned above. Among many others, some of thecomputer programs
currently used are CSANDSET for settlement of shallow footings on sands (U.S. Army
Corps of Engineers); PCStabl for slope stability analysis (Purdue University);
CWALSHT for sheet pile wall analysis (U.S. Army Corps of Engineers); SPT 97 for
capacities of driven piles (Florida Department of Transportation and University of
4
Florida); Shaft 98 for capacities of drilled shafts (Florida Department of Transportation
and University of Florida). The input parameters required for the computation vary from
one program to another. Some require the input of laboratory-test–based parameters
(e.g., PCStabl), while others utilize insitu-test–based parameters (e.g., SPT 97). The
relevant input soil parameters include both strength and stiffness. Strength parameters
determine the limit state beyond which the soil has failed: mainly the cohesion c, the
undrained cohesion su, friction angle φ, or the pressuremeter limit pressure pl. Stiffness
parameters describe the relationship between the applied stress (or imposed
displacement) and the resulting deformation (or induced stress), and are primarily the
elastic modulus E and Poisson’s ratio ν. In addition to the soil properties, knowledge of
the following factors is also as critical in solving geotechnical problems: associated stress
paths, initial stress state, condition of drainage (drained or undrained), and type of
loading (an unloading case or a loading case). In more complicated cases, for instance
when two or more scenarios are working together, the empirical equations or
equilibrium-based equations of the traditional methods prove inadequate to solve the
problem. Then an alternative approach is developed from the boundary value problems
which uses the constitutive models and the finite element methods. The finite element
method applied to the geotechnical engineering is the physical discretization of the
“continuous” soil mass, and approximation by discretization of the mathematical
formulation, which describes the rheological behavior of each element, under a given
type of loading, usually by differential equations or constitutive equations.
Historically, the origin of finite element method can be attributed to the names
such as Hrenikoff in 1941, McHenry in 1943, and Newmark in 1949, mainly for solving
5
problems in solid mechanics (Reddy, 1993; Zienkiewicz & Taylor, 1991a). The finite
element method was primarily considered to be an effective approximation of more
difficult purely mathematical solutions. Mathematicians Richardson in 1910, and
Southwell in 1946 contributed to the development of the finite element method. The
most important technique involves the variational methods by Rayleigh and Ritz in late
1870s and Galerkin in 1915. It was not until around 1956 that Argyris and Turner et al.
formally used the finite element method as a rigorous computation tool for solving
engineering problems.
Currently, the finite element method is widely available in geotechnical
engineering, although its use in practicing engineers is somewhat limited, and computers
programs have been also developed. Some of the finite element programs currently used
to solve geotechnical engineering problems are ABAQUS (developed by the Hibbitt,
Karlsson & Sorensen, Inc.); PLASFEM (developed at the University of Florida); ADINA
(the versatile finite element method software developed by ADINA R & D, Inc.); and
PLAXIS (developed at the Technological University of Delft). PLAXIS is the finite element
code used in this study. Geotechnical finite element codes are featured with constitutive
models to simulate the behavior of different soils under different load conditions. The
most common models included are the Linear Elastic, the Mohr-Coulomb, the Drucker-
Prager, and the Cam-Clay. In addition to the strength and stiffness properties, which are
obtained directly from laboratory testing, (triaxial test, oedometer test) or insitu testing;
the initial conditions and boundary conditions are also specified in the programs.
Empirical correlations have to be used in order to obtain the soil properties from insitu
test data. To date, no constitutive models have been directly developed from the insitu
6
properties of the soil such as the blow count N from Standard Penetration Test, the tip
resistance qc from the Cone Penetration Test, the constrained modulus M from the
Marchetti Dilatometr Test, or the pressuremeter modulus EPMT from the Pressuremeter
Test. Use of correlations introduces additional uncertainty to the various computation
methods. The correlations are established from field observations and laboratory
comparisons. Typically, correlations do not include the statistic parameters that assess
the quality of the correlation between two soil properties: the number of datapoints n, the
regression number, r2, and the standard deviation S.D.. Specific suitability with a
computation method, traditional or finite element method still needs to be addresssed.
Summarizing, geotechnical engineers face the challenge of when obtaining the
best results when analyzing geotechnical problems, selecting among various computation
methods, different correlations, and the choice between laboratory tests and insitu tests.
Objectives of Study
Shown in Figure 1-1 are the four paths possible when approaching a geotechnical
problem. Obviously, the best results should be obtained by selecting the right
computation method, combined with the appropriate input parameters derived from the
most compatible correlations. This study is intended to examine the usefulness of soil
properties derived from the most common insitu tests: Standard Penetration Test, SPT,
Cone Penetration Test, CPT, Marchetti Dilatometer Test, DMT, and Presuremeter Test,
PMT. Since the constitutive models used in finite element methods were founded mainly
on laboratory test results (e.g., the triaxial test), the results from finite element
computations naturally agree well with measured results from such tests. However, the
performance of the constitutive models with the insitu-test–derived soil properties
remains to be known. This study evaluates this combination by comparing the results
7
from the traditional methods using the same parameters, on one hand, with the results
from the full-scale test-measured results, on the other.
Figure 1-1. Different paths possible to solve geotechnical problems.
EQUILIBRIUM THEORIES,EMPIRICAL EQUATIONS
CONSTITUTIVEMODELS
LABORATORY TESTINGTXL, SHEAR, OEDO., …
INSITU TESTINGSPT, CPT, DMT, PMT
Deformation based Problems:• SHEET PILE WALLS• SHALLOW FOOTINGS
CORRELATIONS
INPUT PARAMETERS
FEMANALYSIS
FULL-SCALEFIELD TESTS
PREDICTEDRESULTS
CONCLUSION andRECOMMENDATIONS
ACTUALRESULTS
TRADITIONALMETHODS
PREDICTEDRESULTS
FIELDPERFORMANCE
8
Full-scale tests are considered to be the best way to evaluate soil parameters from
insitu tests and to determine the accuracy of the input parameters used in the
computations. This study was limited to deformation-based problems (as representing
design load behavior, not failure), that is, to deflections under loading, unloading, and
dynamic loading of structures. The unloading case investigates lateral deflections,
bending moments and strut or anchor forces in sheet pile walls. The loading case
estimates settlements of shallow footings. The dymamic loading case is a study of the
Statnamic load test. The four insitu tests used were: the Standard Penetration Test (SPT),
Cone Penetration Test (CPT), the Dilatometer Test (DMT), and the Push-in or the Pre-
bored Pressuremeter Test (PMT). The objective of this work was to determine how the
predictions from the traditional methods and from the finite element method compare
with actual field measurement.
Outline of Thesis
This study constitutes the second part of the final report submitted to the Florida
Department of Transportation (FDOT) Report No. RPWO-14, BC-354, UF Contract No.
450472012. The project is entitled “Evaluation of the FEM Engineering parameters from
insitu tests”. The first part deals primarily with “Finite Element Modeling of Florida Soil
with the Push-in Pencel Pressuremeter”, a doctoral thesis by Brian Anderson in 2001 at
the University of Florida. The second part is the validation of the input parameters with
the case histories using the insitu tests SPT, CPT, DMT and PMT.
This dissertation is composed of nine chapters, among which five chapters are
analyses of real case histories and three of which are full-scale field tests carried out by
the University of Florida. A total of ten calculations are performed.
9
• Chapter 1 is the introduction chapter and explains the origin of the study;
• Chapter 2 is a brief description of the various methods of computation: the traditionalmethods and the finite element code PLAXIS along with the consitutive models;
• Chapter 3 presents a complete review of the four insitu tests SPT, CPT, DMT, andPMT: the fundamental ideas, differences, and discusses the correlations that havebeen given in the literature, including the ones selected for the computations in thisthesis;
• Chapter 4 is devoted to the full-scale field tests on a cantilevered sheet pile wall inTampa, Florida;
• Chapter 5 contains four calculations for sheet pile walls: a full scale test inHochstetten, University of Karlsruhe, Germany, an anchored sheet pile wall inHatfield, Great Britain, and two different wall sections, in Rotterdam, Netherlands;
• Chapter 6 deals with the full-scale field test by University of Florida in collaborationwith the Applied Foundation Testing, Inc., and Pile Equipment, Inc., in Green CoveSpring, Florida;
• Chapter 7 includes the computations of settlement of three shallow footings: two fromthe full-scale field test in Texas A&M (1994), and the third from a mat foundation fora Power Generating Station in Utah;
• Chapter 8 deals with a Statnamic load test on the Green Cove Spring footing, and aStatnamic load test on a steel plate performed in Orlando during the Deep FoundationCongress, 2001;
• Chapter 9 presents the conclusions and recommendations.
The final conclusions of this work should determine which computation method
provides better predictions; determine which insitu tests provide the most adequate input
parameters and correlations; determine how the insitu tests should be carried out; and
determine to which type of problem (unloading, loading, or dynamic) it applies best.
10
CHAPTER 2TRADITIONAL METHODS AND FEM CODE PLAXIS
This chapter discusses the characteristics of the computation methods used in this
study. The basic equations for the traditional methods are presented in Appendix A for
reference. For the finite element method, the focus is on the discretization method and
the constitutive models used by the code PLAXIS. The chapter is divided into four major
parts: the traditional methods for sheetpile wall, traditional methods for settlements in
shallow footings, dynamic anlaysis, and the finite element method with PLAXIS.
Sheet Pile Wall Computations with CWALSHT
The traditional method for the unloading case refers to the computer programs
CWALSHT developed by the U.S. Army Corps of Engineers, 1994. CWALSHT is a
Microsoft Windows-based program that has been used for both educational and
professional purposes since 1996. CWALSHT features a design mode and an analysis
mode for any given sheet pile wall in the ground: whether cantilevered, strutted, or
anchored. The design mode determines the minimum depth of penetration and the
section of the wall for a predefined factor of safety. The analysis mode calculates the
factor of safety for a given depth of penetration and section of the wall. In the calculation
process, CWALSHT uses the Free Earth Support method and the Fixed Earth Support
method for the anchored sheet pile walls. The failure surface can be obtained either by
the “sweep search wedge” method or by the “fixed surface wedge” method (see
CWALSHT Help menu). Rowe’s Moment Reduction is also incorporated in the
program. In a real case, the failure surfaces of cohesionless soils are curves before they
11
reach the active and passive states described by Rankine’s theory (Terzaghi, 1967). This
is best seen in small-scale tests, in both active and passive zones, as described by Lambe
& Whitman (1969). The failure surfaces are clearly curved but the theoretical
formulation of such surfaces is difficult. One possible simplification is the combination
of Rankine failure zone and Prandtl failure zone. The Prandtl failure lines consist of
logarithmic spirals, Figure 2-1 (Prandtl, 1921). The theory from which the traditional
methods for sheet pile walls are based can be found in Bowles (1977), U.S. Army Corps
of Engineers (1999), or Saran (1996).
(45 - φ/2)°
(45 + φ/2)°
Figure 2-1. Combined rupture zones of Prandtl and Rankine.
The Rankine theory is based on the Mohr-Coulomb failure law for granular soils
(c = 0). The active and passive coefficients of lateral pressure for sloping backfill as
shown in Figures 2-2 and 2-3 are given by Eqs. (2.1) and (2.2). In Rankine’s theory, the
wall friction δ is a function of the slope of the backfill β and the inclination of the wall α:
δ = f(β,α).
12
( )( ) ( )[ ]βψαφ
βψφδβαβ
+−×+×+××
+×= 2cossin1
sinsincossinsin
aK (2.1)
( )( ) ( )[ ]βψαφ
βψφδβαβ
−−×−×−××
+×= 2cossin1
sinsincossinsin
pK (2.2)
0
τφ
σ
β
pole
pole
ψ
ψ
σσσ σ
(σf,τf)
(σf,τf)
45°- φ 2 (ψ- φ) 2
vertical directiondirection of major principal stress σ1A
45°+ φ 2
ββ
CA CP
ψ + β
direction of minor principal stress σ3A
failure planes: (σf,τf)
vertical directiondirection of minor principal stress σ3P
direction of major principal stress σ1P
(ψ+ φ) 2
2β
sin ψ = sin βsin φ
Figure 2-2. Plastic equilibrium after Rankine's theory. A) Active. B) Passive.
Alternatively, the general coefficients of earth pressure for a homogeneous
backfill (Figure 2-3A) from Coulomb’s theory are given by Eqs. (2.3) and (2.4). The
derivation of the equations from the Coulomb’s theory is presented in Appendix A.
A B
13
22
2
)sin()sin()sin()sin(
1)sin(sin
)(sin
+×−−×+
+×−×
+=
βαδαβφδφ
δαα
φαaK (2.3)
22
2
)sin()sin()sin()sin(
1)sin(sin
)(sin
+×++×+
+×+×
−=
βαδαβφδφ
δαα
φαpK (2.4)
α
δδ
β
γ φ δ
φ
ρα
δ2
δ1
β
γ1 φ1 δ1
γ2 φ2 δ2
Figure 2-3. Coulomb theory of plane failure line. A) Homogeneous. B) Heterogeneous.
The two methods yield identical equations for the case of a homogeneous soil of
friction angle φ, with smooth vertical wall, δ = 0, β = 0°, and α = 90°.
The following assumptions are used in the calculation methods by CWALSHT:
• The cantilevered sheet pile wall is analyzed as a rigid body rotating about a pointsomewhere in its embedded depth in the free earth method.
• Full active pressures and passive pressures are developed regardless of the magnitudeand the direction of the deflection of the wall. This implies that the plasticequilibrium from the Rankine theory is reached, Figure 2-2.
• Coulomb’s theory of a continuous straight failure plane is used because of simplicity.The error caused by this approximation is considered small (Terzaghi, 1967). In thecase of stratified backfill, a broken failure surface is assumed, and the slope of theline is a function of the friction angle of the layer, Figure 2-3 shows the failure line.
A B
14
• For the structural analysis of cantilevered sheet pile walls, the wall is assumed tobehave as an elastic cantilever beam fixed at the bottom and subjected to the netpressure distribution from the external loads (earth pressure, surcharge) (Figure2-4A).
• For the structural analysis of anchored sheet pile walls, the wall is assumed to have nolateral deflection but can have rotation at the point of anchorage. The free earthmethod assumes that the wall is treated as an elastic beam pinned at the anchor andthe bottom of the wall, Figure 2-4B.
Figure 2-4. Modeling sheet pile wall as beams. A) Cantilevered. B) Anchored.
In CWALSHT, the factor of safety can be controlled at three stages during the
input process for the design mode. At Stage 1, the soil in passive and active zones is
assigned to one factor of safety value. At Stage 2, the factor of safety of the left and right
sides of the wall can be assigned differently. At Stage 3, the factor of safety of each layer
of the two sides of the wall can be individually input (U.S. Army Corps of Engineers,
1994).
Input Parameters for CWALSHT
Using the plastic limit equilibrium theories mentioned earlier, CWALSHT
requires only the strength properties of the soil as input: the internal friction angle φ, the
cohesion c or su, and the moist unit weight γm. For the pile wall, only the flexural rigidity
is required. This involves the modulus of elasticity E and the moment of inertia I of the
A B
15
cross-section of the pile. The longitudinal (or axial) response to the loading from the wall
is not taken into account during the calculation process. Table 2-1 shows the input
parameters for design mode or analysis mode for a sheet pile wall problem with
CWALSHT.
Table 2-1. Main input parameters required for CWALSHT
Input Parameter Name Symbol
Soil
Moisture unit weight γm
Friction angle φ
Cohesion (undrained) c (su)
Adhesion ca
Wall friction δFactor of safety 3 levels of FS
Sheet Pile Wall
Elastic modulus E
Moment of inertia I
Output Results from CWALSHT
The output resulting from running CWALSHT can be divided into two categories:
the output file and graphical output. The output file first echoes the input parameters and
then provides the variation with depth of the following four variables: 1) the deflection of
the wall, 2) the internal bending moment distribution, 3) the internal shear force
distribution, and 4) the net pressure acting on the wall. Along with the factor of safety,
the following values are also provided: depth of penetration for design, maximum
deflection, maximum bending moment and the depth corresponding to the maximum
bending moment. Typical graphical outputs from CWALSHT are shown in Figure 2-5.
16
Figure 2-5. Graphical outputs from traditional method with CWALSHT. A) Deflection.B) Moment diagram. C) Shear diagram. D) Net pressure.
Finally, the magnitudes of the input and output parameters when using
CWALSHT program are all considering a unit thickness in the out-of-plane direction
(parallel to the wall).
Stress Paths in Sheet Pile Walls
The stress paths for rigid wall cases—e.g., concrete walls, can be determined as
long as the soil properties of the backfill are known. However, in the case of driven sheet
pile walls, the driving of the pile into the ground causes disturbance and changes both the
initial state of stress and the soil properties, especially in the vicinity of the wall. These
phenomenona were observed in the sheet piles driven at the Hatfield site (Symons et al.,
1989). The driving of the pile overconsolidates the soil around the pile, but quantitative
A B
C D
17
analysis on this matter has not been done to date. Thus, for driven sheet pile, the
simplified stress paths on the MIT1 p-q space, shown in Figure 2-6 for a cohesionless
soil, would need complementary paths before the excavation (after the driving of the pile)
and after the excavation where the Rankine limit equilibrium theory is assumed.
0
α
Kf-line
Kf-line
K0-line
State of Stress at rest (Normally Consolidated)
Stress path in active zone
Stress path in passive zone
p=(σv+σh)/2
q=(σ
v-σh)/
2
Figure 2-6. Stress paths in Rankine theory from undisturbed K0 stress state.
Settlement Calculations with CSANDSET
The CSANDSET program is a MS-DOS-based program developed by the U.S.
Army Corps of Engineers to estimate the settlements of shallow footings on granular
soils. It was particularly selected for the traditional computations in this work because of
the different methods of settlements computations that it features. The methods of
calculations are based either on the elastic theory or empirical equations. The input
parameters are based on correlations mainly from the SPT and CPT insitu tests. The
methods that are primarily used in the present study are: Elastic Theory method,
Schmertmann (1978), D’Appolonia (1968), Oweis (1979), Schultz & Sherif (1973),
1 The MIT p-q space is a triaxial version of the stress invariants p-q space: the horizontal principal stressesare equal in magnitude. As shown in Figure 2-6, the meaning of p and q are different in the two spaces.
18
Terzaghi (1948), Peck & Bazaraa (1969), Meyerhof (1965), NAVFAC DM-7 (1982).
The origin of each method will be discussed, and as will the assumptions used to derive
them, and finally how they have performed in various predictions reported by other
researchers. The settlement equations for each method are given in Appendix A. The
conventional methods from the DMT and PMT insitu tests will be discussed separately,
after CSANDSET.
Elastic Method
The theory of elasticity may be applied to foundation engineering to calculate
settlements on sand. It is assumed that the footing is flexible and seating on an elastic
half-space (Timonshenko & Goodier, 1970). The basic equation is
∫ ∑=
∆×∆
=⋅=H n
ii
i
i HE
dh0 1
σερ (2.5)
where ε is the elastic deformaion in the layer of thickness H, and ∆Hi the thickness of
sublayers when H is dicretized. In the settlement Eq. (A.7), the Elastic modulus is
estimated in CSANDSET using the SPT blow count N—e.g., Eq. (3.18) in Chapter 3
(Bowles, 1996). Poisson’s ratio is usually taken as ν = 0.3 for sands. It is mentioned in
Bowles (1996) that the Elastic theory method using Eq. (2.5) does not give a good
prediction of the settlements. Some of the reasons are the selection of the parameters
involved in the equation relating to the actual conditions of the footing site, (Bowles,
1996). Nevertheless, in the Elastic theory method, it is assumed that the soil foundation
is homogeneous, isotropic, and elastic.
Schmertmann Method (1978)
The Schmertmann method originated from Elastic theory and is also derived from
the Eq. (2.5). The final form, Eq. (A.8) was obtained from the following approach:
19
( )z
LHVVER I
Eq
qqEq
dzd
×∆
=
∆
∆+∆×−
∆∆
×∆
==σσ
νσ
ερ (2.6)
where the Young’s Modulus E is correlated with the tip bearing from the electric CPT—
e.g., Coduto (1994). The influence factor Iz is a function of the depth and the geometry of
the footing either plane strain (continuous) or axisymmetric (square and circular). The
Schmertmann method is based primarily on the insitu CPT data. Schmertmann method
divides the foundation soil into sublayers, each one having its own soil properties.
Schmertmann’s correction factors C1 and C2, correct for embedment and creep,
respectively. Numerous cases have shown that Schmertmann’s method conservatively
overpredicts the settlement. Gifford et al. (1987) observed from the study of 21 bridge
footings on medium to dense sands using five settlement methods2, that the Schmertmann
method produces larger settlement and has large values of arithmetic mean and standard
deviation. The conservatism of this method is probably due to the underprediction of
stiffness from a strength (bearing) measurement.
D’Appolonia Method (1968)
The D’Appolonia method is an elastic theory-based method as well. The
difference between the settlement calculations with D’Appolonia, Eq. (A.12) and those
with Elastic Theory method, Eq. (A.7) is the correction factors. This seemingly small
difference cause fairly large changes in the settlement prediction. The correction factors
µ0 and µ1 depend on the geometry of the footing. The compressibility of the sand, both
normally consolidated and overconsolidated, is determined empirically as a function of
the SPT blow count N (D’Appolonia et al., 1968).
2 The other four methods are: Burland & Burbidge, D’Appolonia, Peck & Bazaraa, and Hough.
20
Statistical analysis of the settlement predictions of 71 bridge footings showed that
the D’Appolonia’s method generally underestimates the settlements but has a standard
deviation of 1.43: below the average (1.70) from five methods used. The D’Appolonia’s
method provided the most accurate predictions in the case of the 21 bridge footings
mentioned above (Gifford et al., 1987).
Oweis Method (1979)
Proposed by Oweis in 1979, this method is based on the elastic theory, however
the modulus of deformation involved in the computation is dependent on the mean
effective normal stress and the strain level, a nonlinear stress-strain behavior is assumed
(Oweis, 1979). Equation (A.13) shows that the soil mass is divided into n sublayers, each
with an initial elastic modulus estimated from the SPT blow count N. The Oweis method
applies for normally consolidated sands or overconsolidated sands; and for rigid or
flexible footings. As far as accuracy is concerned, the Oweis method gives good
predictions based on the three test cases that Oweis presented when he first proposed this
method in 1979. At present, the Oweis method is considered too conservative. Results
from the settlement predictions for the 5 footings in Texas A & M confirmed this
conclusion (Briaud, 1994).
Schultz & Sherif Method (1973)
Schultz & Sherif (1973) mentioned in their work that this method was initially
derived from the elastic theory, but then a statistical study of 48 measurements on actual
footings and plates allowed an estimate of the soil elastic modulus as a function of the
SPT blow count N. Thus, the settlement equation, Eq. (A.14) is a semi-empirical
equation. Based on the settlement predictions for 71 bridges and the Texas A&M
footings, the Schultz & Sherif method predicts the settlement with good accuracy.
21
Statistical analysis revealed a rather low standard deviation in the 71 bridges (S.D. = 0.64
with average S.D. = 1.7). This affirmation was also true for the case of the Texas A&M
footings (Briaud, 1994).
Terzaghi (1948), Peck & Bazaraa (1969), and Meyerhof (1965) Methods
The settlement equations of these three methods are of the same form. In fact,
both Peck and Bazaraa, Eq. (A.16) and Meyerhof, Eq. (A.17) are a modification of the
equation by Terzaghi, Eq (A.15). The main differences are the correction factors and the
SPT blow count N. Terzaghi’s settlement equation (A.15) is entirely an empirical
equation. The SPT blow count N is not corrected for the Terzaghi method whereas for the
case of Peck and Bazaraa and Meyerhof it should be the overburden corrected blow count
NB. The Meyerhof method does not take into account the influence of ground water.
Looking at the coefficient corrections, Terzaghi’s method usually yields excessive
settlements, which are larger than those from Meyerhof and Peck and Bazaraa.
NAVFAC DM-7 Method (1982)
The NAVFAC DM-7 (1982) method is a more recent method; it is also restricted
only for shallow footings with depth of embedment less than the width B. The modulus
of vertical subgrade reaction Kv1 in Eq. (A.18) is estimated from the relative density, this
latter being correlated to SPT or CPT data (NAVFAC DM-7, 1982: “Soil Mechanics”
Design Manual 7.1). The influence of ground water table is also taken into account by
this method through the vertical subgrade reaction modulus Kv1. Using an empirical
equation, and based on the author’s experience, the NAVFAC DM-7 method generally
overpredicts the settlement, however it is considered to be less conservative than
Terzaghi’s method and Oweis’ method.
22
Input for CSANDSET
Summarizing, the input parameters required for the CSANDSET program are
listed in Table 2-2.
Table 2-2. Main input parameters required for CSANDSET
Input Parameter Name Symbol
Footing dimensions B, L, and D (embedment)
Contact pressure P or q
SPT blow count over depth B N
CPT end bearing over depth B qc
Coefficient of lateral stress at rest K0
Depth of rigid layer H
Depth of ground water table W
Output Results from CSANDSET
The output results from CSANDSET are simply a list of the settlement methods
accompanied by the magnitudes of the calculated settlements. A statistical analysis is also
provided for the different methods: arithmetic mean, median, maximum, and minimum.
Settlement Calculations Using Insitu Tests
The insitu tests that computes settlement of shallow footings directly from the
insitu data are the Pressuremeter Test: PMT (Ménard type) and the Dilatometer Test
DMT. The common point the two insitu tests have is that they measure the strength and
the stiffness properties of the soil by deformation of the soil rather than by failure of the
soil as in the cases of the SPT and CPT.
23
Pressuremeter Method of Settlement Calculation
The PMT method was initially a method proposed by the Centre d’Etudes Ménard
in 1967 (Baguelin et al., 1978). This method is considered to be radically different from
other traditional methods in that it is based on the shear modulus of the soil measured
directly from the PMT (measurement of load-deformation response in horizontal
direction). The soil can be heterogeneous with different insitu elastic moduli. A linear
elastic soil is also assumed, however the settlement equation, Eq. (A.19) shows that the
total settlement is composed of the distortion of the deeper layers and the consolidation of
the layers immediately below the base of the footing. The following term of Eq. (A.19)
BqE c
c××× λα *
9
is for the consolidation part whereas
α
λ
××××
00*
92
BBBq
E dd
is for the distortion part. The PMT method was evaluated first by Dauvisis and Ménard
(1964) for 6 square footings on sand (B = 0.9m), which then yielded to a more empirical
form of the equation, Eq. (A.19). A later study, evaluated 45 footings at 26 sites where
the comparison of the measured settlements and predicted settlements with PMT method
resulted in average ratio close to 1 with the precision within 20 to 30% (Baguelin et al.,
1978). An Excel spreadsheet for calculating the settlement with the PMT method has
been developed by the author and is presented in Appendix A, Figure A-3.
Dilatometer Method of Settlement Calculation
The Dilatometer method estimates the shallow footing settlement on sands by
using the tangent constrained modulus M, Eq. (A.22), which is obtained directly from the
24
DMT. The heterogeneity of the soil is also taken into account as the soil properties are
provided for every 200 mm thick sublayer in the DMT. Equation (A.22) was derived
from the one-dimensional elasticity theory. As to the accuracy, five sites were
investigated by (Hayes, 1986) and sixteen sites were investigated by Schmertmann
(1986). Both researchers concluded that the DMT method provided reasonably good
predictions either from the Ordinary Method or from the Special Method. Computer
programs and calculations spredsheets have been written to perform the settlement
calculations, for instance the Dillyset BASIC program developed by Marchetti (GPE,
Inc., 1993).
Stress Paths in Shallow Footings
In the case of the loading of shallow footings, the initial state of stress is known.
There are no disturbances caused by the installation of structure as in the case of driving a
sheet pile wall or a foundation pile.
0
Kf-line
Kf-line
K0-line
Stress path in active zone
Stress path for point R
p=(σv+σh)/2
q=(σ
v-σh)/
2
0
q=(σ
v-σh)/
2
p=(σv+σh)/2
Kf-line
K0-line
Kf-line
∆qs
OO
M
N
L
M
L
N
I II
R S
Stress path for point S
Stress path for point R
Stress path for point S
L'
L"
Figure 2-7. Stress paths during loading of shallow footing. A) Loose sands. B) Densesands. (Reprinted by permission of John Wiley & Sons, Inc., from Lambe, T.W., Whitman, R. V., Soil Mechanics, John Wiley & Sons, Inc., New York,1969)
A B
25
Figure 2-7 above displays the stress paths during a shallow footing loading for
two cases at two different points in the ground: loose sands and dense sands. Each stress
path starts on the K0 line and eventually reaches failure on line Kf-line. One can notice
the significant difference in the stress increment required in order to reach failure in loose
sands and in dense sands on one hand; and at the point R below the footing and outside
the footing, on the other. A more detailed explanation of the stress paths is provided by
Lambe & Whitman (1969).
Statnamic Test and Analysis with SAW R4
The Statnamic load test is a lon duration impulse test with direct correlation to
static loading. Originally developed in Canada and the Netherlands around 1988 for axial
load test on piles, the Statnamic test has since grown rapidly and become widely used in
many parts of the world. Such growth was marked by the use of Statnamic lateral load
testing in North Carolina in 1994 and Mississippi in 1998, with the capability of reaching
forces up to 7 MN. Under contributions of several universities, and private and
government firms, such as, University of South Florida, Auburn University, Applied
Foundations Testing, Inc., Federal Highway Administration, Hayward Baker, Inc. and so
on, the Statnamic load test keeps expanding with new fields of application: shallow
footings, lateral load test on pile groups. At present, the University of South Florida uses
a 4 MN hydraulic catching mechanism. Such equipment has the advantages of long
duration dynamic loading (~0.5sec), fast mobilization of equipment, and the possibility of
multiple cycles in a relatively short period of time (Mullins et al., 1998). For shallow
footings on sands, the Statnamic test has shown promising and consistent results in
estimating the capacities of the foundations (Mullins et al., 1998). The Statnamic test is
considered as an alternative to the static load tests, which are considerably more time
26
consuming. Details of the procedure for performing a Statnamic load test are provided in
Chapter 8.
Richart (1970) classified the Statnamic test as a “pulse load” in his book
“Vibrations of Soils and Foundations”. The variation of the actual applied load with time
must be measured to analyze the soil response. The other important parameter in a
Statnamic test is the displacement-time relation, or alternatively the acceleration-time
relation. At present, sophisticated devices are used to record displacements (e.g., laser
based device) or accelerations (accelerometer) versus time. Richart (1970) also reported
that when studying the effect of vibrations of soil foundations under a circular base plate
at the DEGEBO3 in Berlin between 1928 and 1936, the experimenters realized that a
mass of soil moved with the footing during the dynamic loading. This idea is also
applied in solving Statnamic problems with the Unloading Point Method. The footing
and the in-phase soil mass are lumped into one mass in the fundamental equations of
dynamic (Lysmer, 1965). Subsequently, in the Unloading Point Method, the foundation
system can be represented by a mass-spring-dashpot as sketched in Figure 2-8.
Figure 2-8. Modeling of Statnamic load testing.
3 DEGEBO: Deutschen Forschungsgesellschaft für Bodenmechanik.
equivalent mass m
Equivalent Damping
Equivalent Spring
Statnamic Loading
[K]
[C]
footing mass mf
Statnamic Loading
in-phase soil mass ms
27
Newton’s Second Law for a single-degree of freedom foundation system is
( )tFuKdtduC
dtudm stn=×+×+×2
2
(2.7)
The fundamental Eq. (2.7) is the basis of the Unloading Point Method (Middendorp et al.,
1992). The UPM is a dynamic analysis adapted to the Statnamic problems. The theory
and assumptions used for this method are discussed in the following section.
Unloading Point Method
The Unloading Point Method uses the simple Kelvin-Voigt model (Malvern,
1969) for viscoplastic materials; it is assumed that the shallow footing combined with the
in-phase soil mass is a concentrated rigid mass m supported by a nonlinear spring and a
linear dashpot in parallel, Figure 2-8. The soil reaction is composed of the static force
and the damping force; the Unloading Point Method is composed of two major steps:
(1) determination of the static load at maximum displacement; and (2) determination of
the mean damping factor Cmean, and subsequently the derived static load-displacement
curve. Referring to Eq. (2.7) and the details of the theory in Appendix A, the static soil
resistance force, Fu(tumax) at the maximum displacement is given by
( )unl
unluu dtudmFtF
×−=
2
2
max (2.8)
where Funl = Fstn(tumax) is the Statnamic load at time of maximum displacement, and
)( max2
2
2
2
uunl
tdt
uddt
ud=
is the acceleration at time of maximum displacement. Knowing
Fu(tumax), and assuming that the soil is yielding over the range Fstn(max) to Funl (Appendix
A, Figure A-4)
unlu FF = (2.9)
28
the damping factors C over that range can be calculated using Eq. (2.10).
dtdu
dtudmFF
Cunlstn 2
2
×−−= (2.10)
and the derived static force Fu is obtained by substituting for C for the mean value Cmean
in Eq. (2.7):
2
2
dtudm
dtduCFF meanstnu ×−×−= (2.11)
The derived static load-displacement diagram can then be plotted as shown in Figure A-4.
Spreadsheet SAW R4
SAW R4 is a copyrighted Excel spreadsheet based on the UPM to analyze
Statnamic problems. The SAW R4 was developed by E. J. Garbin at the University of
South Florida in 2001. The incorporation of the Visual Basic in the program contributes
greatly to the user-friendliness of SAW R4.
Figure 2-9. Input parameters for Statnamic analysis with SAW R4.
29
The main task that SAW R4 does for the geotechnical engineer is the computation
of the mean damping factor, and the plot of the load-displacement curves. Detailed
information on SAW R4 is provided on the website: http://www.sawr4.com. Some of the
characteristic features in the SAW R4 workbook are shown in Figures 2-9 through 2-12.
The input data are imported from ASCII format files such as the *.00X files (X ia
an integer that designates the cycle number), Excel files, etc. The output data can be
exported for a customized plot of the desired parameters.
Figure 2-10. Kinematical parameters versus time.
Figure 2-11. Data option for Unloading Point Method.
30
Figure 2-12. Statnamic output: Load-Displacement curves.
Finite Element Code: PLAXIS
Overview
PLAXIS is a finite element package for analyzing geotechnical problems including
soil-structure interactions. The finite element code PLAXIS comes with standard and
advanced constitutive models that take into account the different aspects of geotechnical
problems such as nonlinear and plastic properties of soils, drainage conditions,
anisotropy, time dependent behaviors, etc. Initially, PLAXIS was the fruit of research on
finite element analysis conducted by then student Peter Vermeer at the Delft University
of Technology in 1974. Written in FORTRAN at the time, it was intended to provide a
FEM-based design tool for an embankment problem in the Netherlands. The capacity of
the program was very limited as it dealt only with elastic-plastic plane strain calculations
on 6-node triangular elements. The original name, ELPLAST, was changed into PLAXIS
(PLasticity AXISymmetry) in 1981 as a result of the extension of the program
31
capabilitites to solve axisymmetric problems with 15-node triangular elements (Borst and
Vermeer, 1984), and incompressible materials (undrained clays). The soil-structure
interaction was developed through the modeling of a beam by Klaas Bakker (Vermeer’s
student) to study the examples of flexible footings and rafts in1990. Other main
contributions were the works by Paul Bonnier in 1989, Ronald Brinkgreve in 1990, and
Harry van Langan in 1991. Since then, PLAXIS has been developed by collaborations
with other universities (specifically the University of Stuttgart in Germany, and the
University Joseph Fourrier Grenoble in France), civil engineering companies, and
government agencies. The company PLAXIS BV was also founded in 1993 and the
PLAXIS research and commercial packages were released, the PLAXIS version 7.1. At
present, conferences and meetings for PLAXIS users are also held periodically in many
parts of the world (Europe, Asia and North America). PLAXIS includes Professional
version 8, and a Dynamic module, which will be discussed below, and the 3-D tunnel
module. PLAXIS is characterized by its flexible user-interface and the robustness in
calculations. Those are reflected in the pre-processor (input), processor (calculations)
and post-processor (output and curves) units of PLAXIS. The special features of PLAXIS
are discussed next.
Input
The drawing window shown in Figure 2-13 for the geometric inputs allows the
user to draw lines for the continuum (soil, rock, concrete block, …) clusters, beam
elements, tie rods, geotextiles (grouting) and interface elements in a CAD-like drawing
fashion. After the material properties have been assigned to the lines and clusters,
PLAXIS proceeds to the particular feature: the automatic mesh generation of triangular
32
elements. Local refinement of mesh around a point, a line, or inside a cluster can be
performed.
Figure 2-13. Sample input window including soil, beam, tie rods, and geotextiles.
The mesh refinement allows the user to locally control the output gradients in an
interested region or a critical region. A very coarse element can be reduced in size by
factor of 20. The triangular finite elements used by PLAXIS are 6-node elements and 15-
node elements with 3 and 12 Gauss points of integration, respectively. PLAXIS uses the
rectangular coordinates ξ and η along with the auxiliary coordinate ζ = 1 - ξ - η, for the
node corresponding shape functions (Zienkiewicz, 1991a). The 3-node line elements are
combined with the 6-node triangular elements and the 5-node line elements are used with
the 15-node triangular elements. The beams are modeled using the Mindlin’s theory
(Bathe, 1995; Zienkiewicz, 1991b). The Gauss points for the beams are as shown in
Figure 2-14. The geotextiles and the interface elements are modeled similarly, however
33
the stress points are positioned according to the Newton-Cotes integration method,
Figure 2-15.
The initial condition is a part of the input process in PLAXIS. This includes the
initial state of stress and the initial hydrostatic pressure (either by seepage or by phreatic
line). In the case of steady state seepage, total discharge, stream lines, equipotential lines
are provided at the end of the mini-calculation. Finally, deactivation of the structures
(footing, walls, anchors, etc.) if applicable, to determine the initial state of stress is
performed at the initial condition stage.
ξ
η
node stress point
η
ξ
(Reduced integration)
Figure 2-14. Nodes and stress points for beam elements.
stress pointnode (Newton-Cotes integration)
6-node adjacent element 15-node adjacent element
interface element
Figure 2-15. Nodes and stress points for interface and geotextile elements.
34
Calculations
In the calculation stage, PLAXIS requires input of the type of calculation to be
used: (1) plastic calculation, which is the most common uses small deformation theory;
(2) consolidation analysis, that considers the time-dependency behavior of the soil and
the pore water pressure; or (3) updated mesh analysis, which is a plastic type of
calculation but uses the theory of large deformation analysis (Lagrangian formulations,
Malvern, 1969). PLAXIS updates the nonlinear equation in the plasticity calculation by
automatically varying the incremental load step. The load increment is reduced when
PLAXIS is confronted with nonlinear response. This feature, illustrated in Figure 2-16
reduces computation time and maximizes accuracy; it also contributes to the robustness
of the PLAXIS calculations.
Figure 2-16. Increment step size varying with respect to nonlinearity.
Depending on the type of the calculation selected, the following parameters in
each phase can be specified: additional step, maximum iteration for the load steps,
35
tolerance error, over-relaxation, arc length control, and load multiplier (plastic type of
calculation); and elapsed time and minimum pore pressure (consolidation type of
calculation), Figure 2-17.
Figure 2-17. Sample calculation set-up with PLAXIS.
The staged construction is a type of loading that is a means for deactivating,
activating and/or changing the properties of any element of the soil and/or structure in a
separate window. Finally, selection of interested points is recommended by PLAXIS
before running the actual calculation. Nodes are selected for displacements, loads, excess
pore pressures whereas stress points for strains and stresses. PLAXIS keeps the user
informed on the above calculation parameters mentioned during the calculation process
(Figure 2-18).
36
Figure 2-18. Calculation process and calculation parameters.
Output and Curves
Graphical output of the calculation results is among the special features of
PLAXIS. By opening the output unit (Figure 2-19) the deformed mesh at the end of the
selected phase (calculation) is first plotted with the deformations in a scale that can be
adjusted by the user (auto scale, manual scale, or true scale).
The following output parameters are available from the user window shown in
Figure 2-19: principal effective and total stresses, displacements (vertical, horizontal and
total), total strains, pore pressure, incremental strains, displacement (useful for locating
critical failure suface during a slope stability analysis), groundwater head, and flow field.
The numerical values of the graphical output data displayed in the window are also
provided in tabular form.
37
Figure 2-19. Deformed mesh with deformations scaled at 50 times larger.
Deformation and stress distributions in any cross section (or multiple cross
sections of the region are obtained by clicking A—A toolbar on Figure 2-19. The
locations of the plastic points (yield function f = 0), plastic cap points (point on the cap
during isotropic hardening) and hardening points, and the tension cut-off points in the
region can be identified as well, as shown in the example of Figure 2-20.
Many possible combinations of the output variables can be plotted in a 2-D
diagram referring to the points that the user preselected before the calculation process
(nodes or stress points). Thus, plots such as, stress paths (Cambridge space), excess pore
pressure versus time, vertical strain versus horizontal strain, bending moment versus load
etc. are provided as well as the numerical values corresponding to the plot (table). The
interactive window for this feature is shown in Figure 2-21.
38
Figure 2-20. Plastic, hardening, tension cut-off points in PLAXIS output.
Figure 2-21. Example of possible 2-D plots in PLAXIS curves.
39
Constitutive Models
PLAXIS simulates behavior of materials with either standard or advanced
constitutive models. The constitutive models are the Linear Elastic model, for rock and
concrete or soil in dynamic problems, the Mohr-Coulomb model, the Hardening Soil
model, and the Soft-Soil-Creep model (PLAXIS version of the Cam Clay model) to model
the viscous and time dependent behavior of the soil. Only the Mohr-Coulomb model and
the Hardening Soil model are discussed in this section, as they are used in this work. The
Linear Elastic model for soil is used in the Statnamic modeling too. The PLAXIS isotropic
linear elastic model obeys Hooke’s Law (Malvern, 1969).
Mohr-Coulomb model
The Mohr-Coulomb model is an elastic-perfectly-plastic type of constitutive
model. It is considered more appropriate as a soil mass model than the Drucker-Prager
model, Fig 2-22 (difference of yield surfaces on the π plane—e.g., modeling of CTE and
CTC4) (Lubliner, 1990). The Mohr-Coulomb model has been used in engineering
practice for modeling most loose to dense sands and normally to over-consolidated sands.
However PLAXIS considers the Mohr-Coulomb model to be only a first-order
approximation of soil behavior. The Mohr-Coulomb numerical solution was improved in
PLAXIS by introducing the potential function in the plastic strain as a function of the
dilatancy angle ψ. The general form of the yield function f is expressed in terms of the
stress invariants I1 and 2J as shown in Eq. (2.12), for a soil with cohesion c and
friction angle φ.
4 CTE: Isotropically Consolidated Triaxial Extension; CTC: Isotropically Consolidated TriaxialCompression.
40
φθθφ
φθθφ
sinsincos3cos3
sinsin3cos3sin
12×−
×+×
×−+=
cIJf (2.12)
The non-associated flow rule in Mohr-Coulomb model by PLAXIS, is written in terms of
the principal effective stresses σ’i (triaxial test) as
φφσσσσ
φφσσσσ
φφσσσσ
cossin21
21
cossin21
21
cossin21
21
21213
12132
32321
×−×′+′+′−′=
×−×′+′+′−′=
×−×′+′+′−′=
cf
cf
cf
(2.13)
for the yield function and fi (fixed yield surface) and
ψψσσσσ
ψψσσσσ
ψψσσσσ
cossin21
21
cossin21
21
cossin21
21
21213
12132
32321
×−×′+′+′−′=
×−×′+′+′−′=
×−×′+′+′−′=
cg
cg
cg
(2.14)
for the potential function gi, where ψ is the dilatancy angle. The flow rule for plastic
strains in case of plane strain problem on the (x,y) plane are:
ijijij
pij
gggσ
λσ
λσ
λε′∂
∂×+
′∂∂
×+′∂
∂×= 3
22
21
1& (2.15)
where i and j = x, y, xy, and yx (λk, k = 1,3 being the magnitude multiplier to the direction
of the strains).
Finite element analyses of strip footings (plane strain) and circular footings
(axisymmetric) by Borst and Vermeer (1984) proved that this Mohr-Coulomb model is
parametrically stable and accurate when predicting the yield stress in the ground. The
non-associative flow rule was combined with 15-node triangular elements to model a
high friction angle foundation soil (ψ > 0).
41
σ'1
σ'2σ'3
θ = - 30°
θ =
+ 30
°
θ =
0°
(CTE)
(CTC
)
CTC: Consolidated Triaxial CompressionCTE: Consolidated Triaxial Extension
Mohr-Coulomb Drucker-Prager
σ'1
σ'2
σ'3
Hydrostatic Axis
fi= 0 (fixed surface)
fi < 0
π-plane
Eref
1
σ'
ε
fi = 0
fi < 0 (Elastic)
Figure 2-22. Mohr-Coulomb model (c = 0): fixed failure envelope and π-plane.
For soil input parameters, PLAXIS Mohr-Coulomb model uses the basic
geotechnical parameters listed in Table 2-3.
Table 2-3. Main input parameters required for Mohr-Coulomb model
Input Parameter Name Symbol
Reference Young’s modulus Eref
Shear modulus (alternative to Eref) G
Poisson’s ratio ν
Effective friction angle φ
Effective cohesion c
Dilatancy angle ψ
42
The reference Modulus Eref can be the initial modulus E0, the secant modulus at
50% of yield stress E50, or the unload reload modulus Eur, depending on the type of the
problem. Apart from the estimation of the elastic modulus from the triaxial test using the
stress-strain plot, laboratory test-based correlations for sands have been established by
authors such as Janbu (1963), Von Soos (1990), Vermeer & Schanz, and NTNU,
“Computational Geotechnics” (Short course on Πλαξισ, Colorado 1999). The most
common equations are given below:
( ) refx
ref ppE σ ′
= 150 to100 for loose sands (2.16)
( ) refx
ref ppE σ ′
= 600 to500 for dense sands (2.17)
where a reference pressure pref = 100kPa in case of SI unit system; and σ’x is the triaxial
confining pressure. The unload-reload modulus can be taken as
Eur = (3 to 5)×E50 (2.18)
Poisson’s ratio generally does not have a significant effect on the analyses (numerical and
geotechnical) as mentioned in Lambe & Whitman (1969). Nevertheless, generally higher
values are used for total stress analyses, 0.30 to 0.45 and lower values, 0.20 to 0.40, for
effective stress analyses. It is customary to use ν = 1/3 for drained conditions and ν =
0.49 for undrained clays; in an unloading case ν ≈ 0.2. The friction angle φ is the
strength parameter of the soil, but it has also influence on the load deformation response
of the soil; high values of the friction angle are not recommended by PLAXIS; the
cohesion c is not very well handled by PLAXIS, in cases of pure sands, inputing small
cohesion (e.g., c > 0.2 kPa) is more appropriate to improve the performance in
43
calculations. Linear variation of the cohesion with the depth and tension cut-off (zero
resistance of the soil in tension) is applicable with Mohr-Coulomb PLAXIS.
Hardening Soil model
The Hardening Soil model is an isotropic hardening model that PLAXIS uses to
model nonlinear behavior of loose sands to dense sands and overconsolidated clays. The
fundamental difference with the Mohr-Coulomb model is the stress-dependency of the
stiffness of the soil and the hyperbolic relationship between the stress and the strain, as
shown in Figure 2-23
asymptote
failure line
axial strain
devi
ator
ic st
ress
|σ1 -
σ3|
qa
qf
ε1
E50
EUR
1
1
Figure 2-23. Hyperbolic deviatoric stress-strain relationship in triaxial test (drained).
The Hardening Soil model describes the hyperbolic stress-strain relationship governed by
the following equation:
for 1
21
501 f
a
qE
<
−
×=ε (2.19)
where qa and qf are functions of the strength parameters c and φ.
( ) 9.0 ; and sin1
sin2cot 3 ==−
×′−×= ff
faf R
Rq
qcqφ
φσφ (2.20)
44
The stiffnesses E50 and Eur that are dependent on the triaxial confining pressure σ’3 and
the strength parameters according to the equations
m
refref
pcc
EE
+×
′+××=
φ
σφ
cotcot 3
5050 (2.21)
m
refrefurur pc
cEE
+×
′+××=
φ
σφ
cotcot 3 (2.22)
where pref = 100kPa; the nonlinearity is represented by the power m which usually takes
the value 0.5 for sands and 1.0 for clays.
The failure envelope for the Hardening Soil model is usually composed of two
different surfaces: a fixed surface, similar to Mohr-Coulomb surface and and a cap
surface which is an ellipsoid in the principal stress space for PLAXIS. The first surface is
the shear surface for the model; it is given by the yield function f in Eq. (2.23)
( )pv
p
ur
a
Eq
qE
f εε −+
−−
×= 150
22
1
1 (2.23)
where εpv is the plastic volumetric strain. The first yield function f indicates that the
failure envelope is not linear as in the case for Mohr-Coulomb when plotted on the (p’-q)
space for m =0.5. The second failure envelope is the expanding cap that describes the
plastic volumetric strain before reaching the failure. It is the oedometer modulus Eoed that
controls the magnitude of plastic strains that originate from the yield cap surface. That
surface is an ellipsoid surface, Figure 2-24, and is given by the equation below
222
2~p
c ppq
f −+=α
(2.24)
45
where p is the mean stress, q~ is a measure of the deviatoric stress, α which depends on
the coefficient of lateral stress at rest K0nc, is the length ratio of the ellipse as illustrated in
Figures 2-24 and 2-25, pp is the preconsolidation related mean stress, obtained from the
initial state of stress and computed from the preoverburden pressure σc’ or the
overconsolidation ratio OCR. The hardening law in the Hardening Soil model with
PLAXIS relates the mean stress pp and the volumetric plastic strain εvpc (on the cap)
1
1
+
×
+=
m
refppc
v p
pm
βε (2.25)
where β is another cap parameter.
elastic region
εvolp
εp
εvolpc
εvolpc
εpc
εpc
pp (initial state of stress) p
q
expanding cap
c cotφ
α pp
f = 0c
f = 0 εij = λpc ∂f c
∂σij
Figure 2-24. Yield surfaces (c ≠ 0) in Hardening Soil in ( p- q~ ) space.
Figure 2-25. Yield surfaces in Hardening Soil (c = 0) in principal stress space.
46
As input parameters for soils, PLAXIS Hardening Soil model uses more basic input
parameters than the Mohr-Coulomb model; those are listed in Table 2-4.
It is customary to estimate the magnitude of the unload-reload stiffness, Erefur, as 3
to 5 times the secant stiffness from the standard drained triaxial test, Eref50. PLAXIS uses
as a default the tangent stiffness from oedometer loading, Erefoed, as identical to Eref
50.
Other insitu-based relations for sands have also been provided by authors such as,
Vermeer or Sandven (1990), and Computational Geotechnics, Colorado (1999):
crefoed qE 3) to(1 = or ( ) ref
crefoed pqE ×±= 1030 (2.26)
where qc is the CPT end bearing of the soil.
Table 2-4. Main input parameters required for Hardening Soil model
Input Parameter Name Symbol
Stiffness parameters
Secant stiffness in standard drained TXL Eref50
Tangent stiffness for primary Oedometer test Erefoed
Power for stress-level depedency m
Strength parameters
Effective friction angle φ
Effective cohesion c
Dilatancy angle ψ
Advanced parameters
Unload-reload stiffness Erefur
Unload-reload Poisson’s ratio ν
Reference stress pref
Normally consolidated value for K0 K0nc
Initial conditions
Pre-overburden pressure POP
Overconsolidation ratio (alternative to POP) OCR (POP)
47
The value of Erefoed is sensitive to small changes when modeling soft clays with
the Hardening Soil model; in the initial condition, input of the Pre-overburden pressure
POP (or alternatively, the overconsolidation ratio OCR) to describe the initial state of
stress is critical for the analysis as it is directly related to the initial position of the cap
and thus the elastic domain for the soil.
Generally, the Hardening Soil model is considered to be a second order
simulation of the soil behavior; the nonlinear and stress level dependency behavior of the
soil are satisfactorily covered by the different parameters listed in Table 2-4.
PLAXIS for Sheet Pile Wall Analysis
The sheet pile wall problems are analyzed as plane strain. The magnitudes of the input or
output parameters are for a unit length (m or ft) in the direction perpendicular to the plane
of the soil-structure model. Α sheet pile wall is modeled as an elastic, flexible beam
discretized into finite rectangular elements whose nodes and stress points were given in
Figure 2-14. Plastic hinges are available to simulate mechanical connections of the beam
with others (e.g., free rotation). As a structural element in the model, each element node
has 3 degrees of freedom (ux, uy and rotation about z axis). The bending moments and the
axial forces are evaluated from the streses at the stress points. As input parameters for
the flexible beam, PLAXIS requires the flexural stiffness EI and the axial stiffness EA,
where E is the elastic modulus for the sheet pile material, I is the moment of inertia about
the axis parallel to the wall, and A is the cross-sectional area od the pile per unit length of
wall. The equivalent beam thickness, deq is calculated from the two above stiffnesses,
Eq. (2.27). Table 2-5 summarizes the input parameters for sheet pile walls.
EAEIdeq 12 = (2.27)
48
Table 2-5. Main input parameters for sheet pile walls
Input Parameter Name Symbol (Dimensions)
Flexural stiffness EI (Force×Length2/Length)
Axial Stiffness EA (Force/Length)
Equivalent thickness deq (Length)
Poisson’s ratio ν
Secondary structural elements used to reinforce the stability of sheet pile wall,
such as anchors and struts can also be modeled in PLAXIS. Anchors, which usually work
in tension, are modeled as two-node, elastic spring elements. The anchors have to be tied
with existing structures such as walls or grout zones (geotextiles). Pre-stressing of the
springs is performed during the staged construction in the corresponding calculation
phase. The properties listed in Table 2-6 describe the strength and geometry of the
anchors.
Table 2-6. Main input parameters for Anchors and Struts
Input Parameter Name Symbol (Dimensions)
Axial stiffness EA (Force)
Spacing out-of-plane Ls (Length)
Pre-stress force (staged construction) P (Force/Length)
Maximum force Fmax (Force)
Struts have the same characteristics as the anchors except that they usually work
in compression. Struts are modeled with one-node elastic spring elements. The input
49
parameters for the one-node struts are the same as those in Table 2-6. Struts and anchors
can be deactivated and activated during different stages of the calculations.
Concrete grout zones are modeled using the geotextile finite elements described
in Figure 2-15. The only property required in the finite element analysis is the axial
stiffness of the grout body. The grout is usually combined with the anchors to simulate
the anchorage system in anchored sheet pile walls.
Table 2-7. Main input parameters for Grout body (geotextile)
Input Parameter Name Symbol (Dimensions)
Axial stiffness EA (Force/Length)
The properties of the interface between a structure and the soil can also be defined
in sheet pile wall analysis. The interface elements introduced in Figure 2-15 are shown in
Figure 2-26 for a practical case. The interface is represented with a strength reduction
factor Rf, which can be taken as 1 for rigid contact between two different materials. For
drained conditions, 0.67 is a common value for modeling the friction between sands and
steel sheet piles, and 0.5 between clays and the steel piles. The minimum possible value
Rf, = 0.01 would simulate the interface between an undrained clay and a steel pile wall.
Figure 2-13 is a complete illustration of an anchored sheet pile wall problem.
The calculation part of a sheet pile analysis with PLAXIS consists of staged
construction where the following tasks can be achieved: activating walls, anchors, struts,
grout body, and soil clusters (e.g., filling in the backfill side); deactivating of soil clusters
for excavation).
50
Figure 2-26. Example of interface elements between a beam and 6-node soil elements.
A B
C D
Figure 2-27. Graphical outputs for sheet pile walls with PLAXIS. A) Deformed mesh. B)Lateral Deformation. C) Bending moment. D) Shear stress.
51
In addition to the output mentioned in the section “Overview” on PLAXIS, the
output for sheet pile wall analysis consists of the displacements parameters (ux, uy and
rotation about z axis), the bending moment diagrams, the shear forces and the net earth
pressure distribution on the beams. The post-calculation anchor force and the strut force
are also provided, and finally the stress distribution along the interface elements. The
output results can be viewed in plots and in tables. An example of graphical output for a
sheet pile wall is shown in Figure 2-27.
PLAXIS for Shallow Footing Settlements
The finite element analysis with PLAXIS treats circular footing problems as
axisymmetric problems. A square footing can be transformed into an equivalent
axisymmetric problem and strip footings are modeled as plane strain problems.
Prediction of deflections of shallow footings with finite element methods is a rather
simple task for geotechnical engineers. Thus, only few particular points are emphasized
for the case of PLAXIS. The footing structure can be modeled as flexible or rigid. For a
flexible footing, the user models the footing as a flexible beam with the elastic properties
presented previously in Table 2-5. For a rigid footing, the user can simulate the footing
with a nonporous (zero permeability), elastic material; for instance a relatively thick
concrete footing or steel plate footing. The interface between the base of the footing and
the soil has a minor effect on the results. The PLAXIS Manual (1998) shows that the
choice of a rough or smooth footing can make a difference of about 2% in the estimation
of the failure load.
PLAXIS for Dynamic Analysis
The dynamic module of PLAXIS may be used to analyze a Statnamic load test
performed on a shallow footing. As a quick review of dynamic problems in geotechnical
52
engineering, the soil and structures in the finite element model obey the fundamental
equation in dynamic problems, Eq. (2.28)
[ ]{ } [ ]{ } [ ]{ } { }FuKuCuM =++ &&& (2.28)
where {u} is the displacement vector; { },u& the velocity vector {ü}; the acceleration vector;
[M], the mass matrix; [C], the damping matrix; [K], the stiffness matrix; and {F}, the
load vector. The damping matrix [C] in Eq. (2.28) represents the viscosity of the
material; it is written as a combination of the mass matrix [M] and the stifness matrix [K]
(Zienkiewicz, 1991b):
[ ] [ ] [ ]KMC RR ×+×= βα (2.29)
such that αR and βR are called the Rayleigh damping coefficients.
Both plane strain and axisymmetric models may be loaded in PLAXIS with the
three types of dynamic sources: uniform harmonic loading, Eq. (2.30) (e.g., machine
vibration).
( ) ftFF ×=+××= πωϕω 2 with sin 0max (2.30)
a load multiplier input from an ASCII or SMC file (e.g., earthquake loading) or block
loads, which can also be a special case of harmonic loading or downloaded from an
ASCII file (e.g. the driving of a concrete pile). The major differences between the static
and dynamic finite element analyses are in the input and calculation processes. In the
input process, the geometry of the problem has to be large enough to avoid reflection
disturbances from the boundaries. For instance, in the case of a circular shallow footing
of diameter 2R, the approximate domain dimensions required are presented in Table 2-8.
In addition to the size of the domain, absorbent boundaries are also added to minimize the
increments of stresses on the boundary caused by the dynamic loading.
53
Table 2-8. Difference in geometry model between static and dynamic problems
Parameters Dynamic Domain Static Domain
Diameter of shallow footing (2R) (2R)
Width 40R 5R
Depth 20R 8R
The finite element mesh is generally coarse for points far from the structures.
However, around the structure, more refined meshing is recommended, as is the case for
the static problem. A typical input window is presented in Figure 2-28 where the
absorbent boundaries are represented by the thick boundary lines at the bottom and the
right side of the domain.
Figure 2-28. Example of geometry input for dynamic analysis.
The soil is usually modeled as an elastic material, especially for small
displacement problems (vibrations, Statnamic simulation, …). However, for cases such
54
as pile driving, the Mohr-Coulomb model or the Hardening Soil model may be more
appropriate to use. In addition to the soil properties mentioned in the previous
paragraphs, the shear wave velocity is required to characterize the dynamic
compressibility of the soil. The equation for the shear wave velocity Vs can be found in
many references (Lambe & Whitman, 1969; Richart, 1970; Das, 1983) and is repeated
below for convenience.
( )νρ +==
12 and EGGVS (2.31)
where G is the shear modulus and ρ = γ/g the mass density of the soil.
The Rayleigh damping coefficients αR and βR are input in the program during the
calculation process. For the time integration step, the stability and accuracy of the
calculation process in PLAXIS are based on the Newmark scheme developed by Sluys
(1992). They are, for the displacement and the velocity, respectively
2
21 tuutuuu ttttttt ∆×
×+×
−+∆×+= ∆+∆+ &&&&& αα (2.32)
( )[ ] tuuuu tttttt ∆××+×−+= ∆+∆+ &&&&&& ββ1 (2.33)
with the condition for stability of calculation
5.0≥β and 2
21
41
+≥ βα (2.34)
and the critical time step for a single element is provided in the work by Pal (1998).
Besides the static output parameters, the following standard outputs (graphical,
plots, and tables) is provided in the dynamic analysis: Acceleration, Velocity, and
Displacement versus Time parameter. A sample output curve is shown in Figure 2-29.
55
Figure 2-29. Pile tip displacement versus time during driving.
56
CHAPTER 3LITERATURE REVIEW FOR INSITU TESTING
General
By definition, insitu tests are geotechnical tests performed in the ground to
determine the physical and mechanical properties of the soil. Some engineers consider
insitu testing as an alternative or complement to laboratory tests in determination of the
classic soil properties such as: friction angle of sand φ’, undrained cohesion of clay su,
coefficient of lateral pressure at rest K0, shear modulus G, etc. Other engineers consider
sample disturbance so important that they may rely entirely on insitu testing. The
fundamental difference between the two test methods is that insitu tests often require
empirical correlations to determine the properties mentioned above, whereas laboratory
tests usually provide them directly. A brief comparison of laboratory and insitu testing is
presented in Table 3-1.
Table 3-1. Laboratory testing versus Insitu testing
Laboratory Testing versus Insitu Testing
Advantages
Can carefully control stress path anddrainage condition Test under insitu conditions
Disadvantages
Disturbance due to sampling,transportation, preparation
No choice of stress path: slow rate:drained; fast rate: undrained
Disturbance due to stress relief Disturbance due to insertion
57
In engineering practice, laboratory and insitu tests can and should be combined
for accurate soil characterization (Lunne et al., 1997). In Florida where the ground water
table is very often shallow and the subsurface soil type is essentially cohesionless, it is
impractical to obtain a good quality sample for laboratory testing. Therefore, insitu tests
may prove more appropriate for a site investigation. The correlations have to be used
carefully as, besides the inconsistency from the variability on the site, uncertainty
resulting from various other factors is introduced (equipment, scattered data, etc). In
each of the insitu tests, there are two or more different methods or types of equipment
available. For example, in the Pressuremeter test, PMT, one may choose between the
pre-bored PMT, the self-boring PMT, or the pushed-in PMT or driven PMT. However,
when the correlations are established to obtain strength and stiffness properties of the
soil, the differences in equipment types and methods are rarely emphasized. On the other
hand, the constitutive models implanted in the FEM programs are mainly calibrated
laboratory testing. To improve the reliability of using insitu tests with finite element
methods, the testing procedures, equipment, and correlations must be studied. Four types
of insitu test are discussed in this chapter, the Standard Penetration Test, SPT; the Cone
Penetration Test, CPT, the Flat Plate Dilatometer Test, DMT; and the Pressuremeter Test,
PMT. The first two tests measure the strength and the stiffness by failing the soil in
vertical direction, whereas the last two measure the strength and the stiffness of the soil in
lateral direction.
Standard Penetration Test: SPT
Background and Procedure
Charles Gow in 1902 initially introduced the SPT as a method for recovering dry
sand samples. The equipment that he used was a 1-inch diameter sampling tube inserted
58
into the ground with a 110-pound hammer (Riggs, 1986). In 1927, L. Hart and G. A.
Fletcher devised the 2-inch split spoon combined with the mass of 140 lb and a 30-inch
drop. Later on, this type of sampler became standardized which is now called the
Standard Penetration Test. Figure 3-1 illustrates the split spoon for the SPT. The
traditional way to drop the 140-lb hammer is to lift the mass with a rope then cause the
free fall by quickly releasing the friction of the rope against the cathead, Figure 3-2.
split tube
Figure 3-1. Type of standard 2-inch diameter split spoon by Hart and Fletcher (1927).
Figure 3-2. Original equipment used to perform SPT. (Reprinted by permission of NIST,from Kovacs, W. D., Salomone, L. A., Yokel, F. Y., Energy Measurementsin the Standard Penetration Test, Building Science Series 135, NationalBureau of Standards, Washington, DC, 1981)
59
The strength of the soil is measured by the number of blows, N, to drive the sampler 12
inches at the bottom of a drilled hole. In 1948 when Terzaghi and Peck first proposed in
their book the correlation between blow count SPT N and the Relative Density Dr of sand
deposits, “Theory and Practice in Soil Mechanics” (1948). Changes in the test procedure
occurred in 1954 when James Parsons proposed counting the blow for each of the last
three 6 inches increments and adding the two lowest values to give the value SPT N
(Palacios, 1977). Nowadays in the U.S.A, the SPT is standardized as ASTM D 1586.
The dimensioned sizes of the ASTM standard split spoon are presented in Figure 3-3.
Figure 3-3. ASTM standard split spoon. (Reprinted by permission of ASTM, fromAmerican Society for Testing and Materials, Annual Book of ASTMStandards, Vol. 04.08 (Soil and Rock; Dimension Stone; Geosynthetics),ASTM, Philadelphia, 1991)
The ASTM procedure consists of recording the number of blow counts in the last two 6-
inches of the 18 inches during the driving of the sampler. The main reasons for this
change from original procedure as found in the work of Palacios (1977) are: disturbance
of the soil in the first 6 inches during the drilling operation at the bottom of the hole;
uncertainty on how clean the hole is, before the sampling process starts; and stress relief
in the top layer below the hole which was considered to extend for few inches below the
bottom of the hole. The idea of the influence of the effective overburden pressure on the
blow count N was originally from Gibbs and Holtz (1957) by conducting laboratory tests
60
and finding the relationship between the three parameters: relative Density Dr, effective
overburden pressure and the blow count N. Schmertmann (1977) stated that besides the
effective overburden pressure, the blow count is also affected by the stress history of the
soil, which was then cofirmed by Marcuson and Bieganousky (1977) when establishing
the equation, Eq. (3.1).
21
0045.0psi)(24.02.34.10
83.06.8
−−++= v
rOCRN
Dσ (3.1)
Today, the SPT is the most widely used insitu test in U.S.A, mainly because it is
quick, simple and inexpensive (Kulhawy and Mayne, 1990). The use of SPT has grown
rapidly, three different types hammer systems have been developed: donut hammer,
safety hammer, and automatic hammer.
The safety hammer system is similar to the donut hammer system in that the
weight is raised and dropped by the operator as shown in Figure 3-2. The main
difference is that the safety hammer reduces the risk of injury (Lamb, 1997). Unlike the
first two types of hammers, the automatic hammer system is a hydraulically powered
chain lift device (Riggs, 1986). The rate of drop can be controlled up to maximum of 50
blows per minute. The only thing that operator does is to open the hydraulic valve in
order to start the lifting and dropping sequences. Apart from the test execution, the
energy transfer from each type of hammer system is also different. This is one of the
factors associated with the different results from SPT. Spoor (1997) provided the
difference in energy ratio from different hammer systems. ASTM Standard D 1586
allows the use of a wide variety of equipment. At this point, it is natural and a fact that
the SPT is a controversial test. The parameter used to measure the energy is the Energy
61
Efficiency Ratio or the Energy Transfer Ratio. Two methods, the Fv (EFV) method and
the F2 (EF2) method, are available to calculate the energy transfer ratio; details can be
found in the derivations of the equation by Butler (1997). Equation (3.2) to Eq. (3.3)
calculate the energy efficiency ratio ERFV and ERF2 from the two methods.
( ) ∫∫ ×=×==max
0
d)()(d)( ttvxFtdtdxxFtEEFV (3.2)
∫×==
2L/c
0
2d)(2 tFAE
ctEEF (3.3)
lb-ft350EFVERFV = (3.4)
lb-ft3502
2EFERF = (3.5)
where, F is the force; v, the velocity; c, the wave velocity in steel; E, the modulus of
elasticity of the steel rod; A, the cross sectional area of the rod; and 350 ft-lb represents
the maximum potential energy of a hammer system. U.S. practice has standardized the
SPT N at 60% of the hammer energy being transferred to the drill rods, (Seed et al.,
1985). This leads to the standard value N60 given by Eq. (3.6)
fieldmeasured NE
EN ×=
6060 (3.6)
where E60 = 0.60×350 ft-lb, is the theoretical potential energy, Emeasured = ERFV or ERF2
the measured energy, and Nfield, the N-value observed in the field. Skempton (1986) took
into account the dimensions of the equipment and converted the raw (measured) SPT N-
value into N60 as in Eq. (3.7)
NCCCE
N RSBm ××××
=60.060 (3.7)
62
where Em is the hammer efficiency based on Clayton (1990), CB, CS, and CR, are the
borehole diameter correction, the sampler correction, and the rod length correction ,
respectively (Skempton,1986). Other approaches have been proposed to estimate the
energy related SPT N, the most recent is the method proposed by Bowles (1996) to
calculate the corrected SPT N at 70% of energy efficiency, Eq. (3.8).
4321
21
070
76.95 ηηηη ×××××
′
=′ Np
N (3.8)
where p’0 (in kPa) is the effective overburden pressure, ηi, are adjustment factors
depending on the type hammer system, sizes of the rod, sampler type and dianeter of the
borehole (Bowles, 1996). Bowles (1996) also provided the equation to convert the SPT
N for any other energy level knowing N’70 from Eq. (3.8).
Summarizing, the SPT is greatly affected by the type and sizes of the equipment
(Schmertmann, 1978). Other factors that can affect the SPT N are: variation from the
exact 30-in. drop of the drive weight, use of deformed tip on the sample spoon (old
equipment), carelessness of the operator (counting the N-value, operating correctly) and
so forth (Kulhawy et al., 1983).
Correlating Soil Properties from SPT Blow Count N
Despite the implied nonreproducibility of the test mentioned above, geotechnical
engineers have established correlations with the SPT N for almost all of the soil
geotechnical properties. As shown in Eq. (3.1), the relative density of sands was among
the first correlated to the SPT N. Now, the physical properties, insitu state of stress,
strength, and stiffness of the soil can be correlated using tables, charts, and/or empirical
63
equations. The correlations are different for cohesionless soils (sand) and cohesive soils
(clay); they are based of the 60% energy efficiency of the hammer system.
Correlations for cohesionless soils
Physical Properties. The evolution of the correlation from the very simple and
direct correlation of the relative density Dr with the SPT N was by Terzaghi and Peck
(1967) or Lambe and Whitman (1969), improved by Gibbs and Holtz (1957), and then by
Marcuson and Bieganousky (1977) were already discussed in the previous section.
5.020 50779711231122275.02.12(%)
−−−++= u
a
vr C
pOCRND
σ (3.9)
where pa is the atmospheric pressure and Cu is the particle size distribution (r2 = 0.77).
The study on SPT calibration data from Japan, China, U.K. and U.S.A by Skempton
(1986) further improved Eq. (3.9), it is then concluded that the relative density Dr is
related with the SPT N as in Eq. (3.10)
NCCC
CCCCCD
OCRAP
NRSBERr ×
××××××
=2 (3.10)
where N is the field measured SPT N-value, and the corrections Ci are obtained from
charts, empirical equations or tables provided in Kulhawy and Mayne (1990).
Estimation of unit weight of cohesionless soils from the SPT N is not common in
the literature. The correlation used in this research was the one found in the Florida
BPier Manual.
Insitu state of stress. The insitu state of stress is quantified with the coefficient
of lateral stress at rest K0 or the overconsolidation ratio OCR; they are defined as follows
0
00
v
hKσσ
= (3.11)
64
0v
cOCRσσ
= (3.12)
where cσ is the preconsolidation stress (named as the preoverburden pressure, POP in
Chapter 2 dealing with Πλαξισ). The lower bound and the upper bound for K0 are the
active pressure coeffient Ka and the passive pressure coefficient Kp, already discussed in
Chapter 2. The most complete description of the stress history when loading a soil is the
paper written by Mayne and Kulhawy in 1982. They showed the possible variation of the
effective horizontal stresss 0hσ and the effective vertical stress 0vσ (therefore change in
value of K0) during a simple loading-unloading-reloading case.
Direct correlations between the SPT N and the insitu state of stress, K0 and OCR,
have not been developed so far. However, it is possible to evaluate the coefficient of
lateral stress K0 indirectly from SPT N using the Cone Pressuremeter test data (as it will
be shown in the next sections, the SPT N and the CPT end bearing qc are related).
Strength parameters. The strength parameters for sands are primarily the Mohr-
Coulomb theory based parameters: effective (stress) friction angle φ and effective (stress)
cohesion c. For dense sands, the friction angle at failure is the peak friction angle φp,
before the soil starts to dilate during a triaxial compression test. Whereas for loose sands,
the friction angle at failure corresponds to the critical void ratio friction angle φcv, which
is considered to be the same as the peak friction angle φp (Kulhawy and Mayne, 1990).
In geotechnical engineering, there are two types of friction angle: plane strain friction
angle φps, and triaxial compression (also called axisymmetric friction angle φax,
Schmertmann, 1988) friction angle φtc. They are empirically related to each other; for
instance, Ladd and Lee (1976) suggested
65
°≤=
°>°−=
34for
34for 175.1
tctcps
tctcps
φφφ
φφφ(3.13)
which is written in another form by Schmertmann (1988)
°≥°−
=
°<=
32for 3
32-
32for
psps
psax
pspsax
φφ
φφ
φφφ(3.14)
The earliest correlations for estimating φtc with the SPT N were established by
Meyerhof (1956) in form of tables. About two decades later, in 1974, Peck, Hanson and
Thornburn, introduced a more conservative estimation of φtc. The Peck et al. (1974)
correlation Eq. (3.15) is used to estimate the friction angle from the insitu SPT in this
research.
( )N0147.0exp6034.27881.53tc −−=φ (3.15)
Schmertmann (1979) provided a chart to estimate φtc from the SPT N. The relationship
includes the overburden stress 0vσ and approximately yields into Eq. (3.16).
34.0
0
1tc
3.202.12tan
+
= −
a
v
p
Nσ
φ (3.16)
The friction angle value from Eq. (3.16) is considered to be too conservative and not
recommended for sands at very shallow depths (< 1.0 to 2.0 m) (Kulhawy and Mayne,
1990).
Stiffness parameters. The nonreproducibility of the SPT is most pronounced in
the correlations between the compressibility, elastic modulus E or constrained modulus
M, of the soil and the SPT N-values. In fact, many correlations have been established
(e.g., Bowles, 1996) and they showed significantly different results (Coduto, 1996;
66
Kulhawy and Mayne, 1990). The type of modulus most used in correlations refers the
secant modulus assuming a drained condition during the triaxial test. Correlations for the
initial modulus Ei and the tangent modulus Et are not common. Kulhawy and Mayne
(1990) proposed as a first order correlation the following equations
sands idatedoverconsol clean,for 15
sands edconsolidatnormally clean,for 10
fines with sandsfor 5
60
60
60
NpE
NpE
NpE
a
a
a
=
=
=
(3.17)
where pa is the atmospheric pressure. Bowles (1996) mentioned that, a good estimation
of the secant modulus from the SPT is
( ) sands edconsolidatnormally for [kPa] 15500 += NE (3.18)
sands idatedoverconsolfor [kPa] NCOCR OCREE ×= (3.19)
The SPT N-value in Eq. (3.18) and Eq. (3.19) is estimated at 55% of energy efficiency.
However, in the practice, engineers usually use the measured N or N60 for simplicity. The
last two correlations are used in this research.
For dynamic problems, the estimation of the dynamic elastic modulus is
calculated from the shear wave velocity Vs. Horikoshi et al (1998) have shown in their
work that the empirical equation, Eq. (3.20) is practical to estimate the shear wave
velocity of the sand from the SPT N.
sands ne)(Pleistoce diluvialfor [m/s] 2.97
sands (Holocene) alluvialfor [m/s] 6.80323.0
331.0
NV
NV
s
s
=
=(3.20)
67
Other correlations based on the SPT N and having the same form as Eq. (3.20) are from
Jamiolokowski (1988) and Seed et al. (1986); they also account for the geological age of
the sand and the considered depth (Bowles, 1996).
Correlations for cohesive soils
Physical properties. The consistency index CI is the parameter that determines
the compactness of cohesive soils. It has the same meaning as the liquidity index LI and
they are related to each other by Eq. (3.21).
LIwwww
CIPL
nL −=−−
= 1 (3.21)
where wn is the natural water content and wL and wP are the liquid limit and the plastic
limit, respectively. The SPT N and the consistency of cohesive soils have been correlated
by Terzaghi and Peck (1967). Szechy and Varga (1978) also provided a quantitative
relationship between the SPT N and the consistency, CI. For example, a very hard clay
of CI > 1.5 corresponds to a blow count N greater than 30.
Insitu state of stress. The insitu state of stress of cohesive soils has been
extensively studied by geotechnical engineers. Many correlations relating the coefficient
of lateral stress K0 or the overconsolidation ratio OCR are available. The accuracy of the
correlation is quantified using the statistical characteristics (n, r2 and S.D.). For the
parameter K0, Kulhawy et al. (1989) provided a chart that includes datapoints of 13 intact
clays and 5 fissured clays. The trendline is Eq. (3.22).
)43.0.. ;771.0( 073.0 2
00 ==×= DSrpNK
v
a
σ(3.22)
68
In terms of overconsolidation ratio, Mayne and Kemper (1986) provided an even
more scattered data correlation. Equation (3.23) describes the trendline with the number
of datapoints n, included.
)82.3.. ;661.0 ;112( 58.0 2
0===×= DSrnpNOCR
v
a
σ(3.23)
The preconsolidation stress cσ is a useful parameter to back-calculate the coefficient of
lateral stress K0. It was correlated with SPT N, by Mayne and Mitchell (1988), Eq.
(3.24).
)37.4.. ;699.0 ;126( 47.0 2aac pDSrnpN ===×=σ (3.24)
Strength parameters. The effective strength parameters φ and c are generally
not used by engineers for cohesive soils in their design. Due to the low value of the
hydraulic conductivity (or permeability), clays are usually assumed to have undrained
behavior. Moreover, the only insitu test that measures both the effective friction angle φ
and the effective cohesion c directly is the Iowa Borehole Shear test (BST). SPT, CPT,
and DMT correlations provide the undrained cohesion su for cohesive soils, and the
effective friction angle φ for cohesionless soils.
Reproduced by Lambe and Whitman (1969), Terzaghi and Peck (1967) first
proposed ranges of values relating the SPT N with the undrained shear strength su.
Correlations from several authors have been collected by Djoenaidi (1985); the rather
significant scatter can be attributed to different energy efficiency, equipments, and
standardisation of the SPT (Kulhawy and Mayne, 1990).
69
More uniform datapoints (from the same equipment type and SPT procedure)
were also found in Kulhawy and Mayne (1990). The scatter is much less and the number
of datapoints is high as they are shown in Eq. (3.25).
)865.0 ;180( 29.0 272.0 =≈= rnNps
a
u (3.25)
Table 7.4 in Lambe and Whitman (1969) (page 77), and Eq. (3.25) are used in this
research to evaluate the undrained cohesion su of clay materials from SPT.
Stiffness parameters. The elastic modulus of cohesive soils can be the
undrained elastic modulus or the secant modulus. Direct correlations between the secant
modulus Es with SPT N are rarely found in the literature. It is more common to relate Es
with the undrained cohesion su of the clay. A very rough relationship between the two
parameters was found in Bowles (1996) and used in this research, Eq. (3.26).
Claysandy or Silty for 1500) to500(silt andClay for 500) to100(
us
us
sEsE
==
(3.26)
The dynamic elastic modulus for cohesive soils is estimated in the same manner
as in cohesionless soils. The correlation is between the SPT N and the shear wave
velocity Vs, and it depends on the geological age of the soil, Eq. (3.27).
clays ne)(Pleistoce diluvialfor [m/s] 114
clays (Holocene) alluvialfor [m/s] 102294.0
292.0
NV
NV
s
s
=
=(3.27)
Advantages and Disadvantages of SPT
The SPT is considered to be the most widely used insitu test in the world. It is
simple to use, economical, and fast. The SPT allows recovery of soil samples from the
investigation site. It is also able to penetrate almost any type of soil: hard layers, gravels,
etc. On the contrary, the SPT, as discussed above, is marked by the nonreproduciblity of
70
the data due to the variety of equipment and/or difference in the test procedure (human
error). The SPT is not useful either when dealing with soft clays because fewer property
correlations were available.
Cone Penetration Test: CPT
Background and Procedure
The first form of cone penetration test was the Dutch cone penetrometer test. It
was made by the engineer P. Barentsen in 1932 at the Rijkwaterstaat, the Department of
Public Works in the Netherlands (Lunne et al., 1997). Barentsen used a gas pipe of
19mm inner diameter, inside which a 15mm steel rod could move freely up and down.
The inner steel rod had a 10cm2 projected area cone with 60° apex angle at its tip. The
test consisted of manually pushing the cone in the ground through the steel rod and
measuring the penetration resistance with a manometer. The Dutch cone penetration test
was then sometimes called the static penetrometer test, deep penetration test, or the quasi-
static penetrometer test. In 1935, T. K. Huizinga devised the first cone penetration rig
and changed the 15mm inner diameter of the pipe to 19mm to reduce the tube-rod friction
during the pushing. In 1928, Verdeimen, and then Plantema added to the original cone a
tapered sleeve in order to keep soil out of the gap formed when the cone (inner rod) alone
is pushed. It was not until 1953 that Begemann improved greatly the Dutch cone by
adding the “friction jacket” behind the cone, to measure the local skin friction in addition
to the tip resistance. It is the Begemann cone penetration version, also called the
mechanical cone penetration test that is still kept until the present day. Several European
countries, such as Germany, Belgium, and France have adopted their own type of
mechanical cone penetration devices (Lunne et al., 1997).
71
A huge improvement was achieved in Germany when the first electric cone
penetrometer was made at the DEGEBO, Berlin during the World War II (Broms and
Flodin, 1988). It was in 1949, when Delft Soil Mechanics Laboratory (DSML) measured
separately the tip resistance and the sleeve friction with the electric cone penetrometer.
The electric cone penetrometer is more sophisticated and is equipped with a load cell,
strain gauges, friction sleeve and other elements.
The electric cone penetrometer test procedure pushes the cone into the ground at a
constante rate of 20mm/s, the data acquisition system at the ground level records the soil
resistance (end bearing and skin friction) instantly transmitted from the sensors. The
ASTM standardized the cone penetration test in 1986 as ASTM D 3441, and allows three
types of cone penetrometer tips: the Dutch mantle tip, the Begemann tip, and the electric
cone tip. The procedures and limitations for each type of cone penetrometer are fully
detailed in ASTM D 3441. The piezometer probes, or simply called piezocone, also
arrived in 1974, when the Norwegian Geotechnical Institute (NGI) developed their
conventional electrical piezometer. Several researchers have contributed towards
developing the piezocone to measure the pore pressure during the cone penetration: Janbu
and Senneset in 1974, Schmertmann in 1974, Torstensson, and Wissa et al. in 1975. The
Wissa type piezocone was then changed by Schmertmann in 1978 to have the
conventional 60° with filters at the tip when he studied the liquefaction potential of sands.
Currently, all of the previously discussed types of cone penetrometer are pushed
into the ground with a hydraulic ram equipped in a “cone truck”. The dead weight of the
truck is usually about 20 tonnes, which is also the maximum thrust capacity of the
equipment in the geotechnical engineering practice. Lunne et al. (1997) display typical
72
interiors and exteriors of various types of existing cone trucks in their book: “Cone
Penetration Testing in Geotechnical Practice”. The University of Florida, geotechnical
Group has had its cone truck since 1985. A full description of the University of Florida
cone truck including the data acquisition process with the special features from the cone
tip to the final output, Figure 3-4, was published by Davidson and Bloomquist (1986).
Besides the CPT, the Pencel type PMT and the Flat DMT are also carried out using the
cone truck.
Figure 3-4. University of Florida cone truck, outside and inside views.
The two basic parameters obtained from the CPT are the tip resistance, denoted
by qc or qt, and the sleeve friction, fs. The friction ratio Rf (%) is given by the ratio in Eq.
(3.28). It is a very important parameter for identifying the soil type.
100×=c
sf q
fR (3.28)
The data acquisistion system reproduces instantly the magnitudes of the tip resistance and
the sleeve friction versus depth, plus the cone penetrometer inclination and the pore
pressures if applicable (Davidson and Bloomquist, 1986). In the case of the electrical
CPT, interval time for the data readings can be as short as corresponding to 10 to 50 mm
of depth of sounding. For instance, with the Hogentogler software “Cptsnd”, which is
73
used at the University of Florida, data readings are taken every 50 mm. The collected
data are copied to a floppy disk and brought into office in order to obtain the soil
identification (physical properties), the strength parameters and the stiffness parameters.
At present, the correlations are incorporated into computer programs, compatible with the
raw data file. The data reduction program used in this study is Cptintr1 v. 3.04.
The mechanical and the electrical CPTs are more versatile sounding procedure.
Established correlations enable identification of soil type and determination of the
engineering properties along the soil profile. Other sensors have been added to the cone
penetrometer as well: geophones for seismic cone test, electrical resistivity, etc.
Disturbance Due to Insertion of Cone Penetrometer
The earliest study on the effect of penetration due to insertion of the probe was by
Baligh and Scott in 1975. In their theoretical and experimental study, Baligh and Scott
(1975) used plane strain rigid wedges with various apex angles to penetrate in clays. It
was not until 1983 when Davidson and Boghrat (1983), using a pseudo stereo photograph
technique at the University of Florida, performed laboratory tests to study the effect of
the insertion of the standard axisymmetric cone penetrometer in loose sands and in dense
sands. Displacement directions and magnitudes, volumetric strains, and shear strains
around the probes were measured. It was observed that the directions of the displacement
of the soil adjacent to the cone are different in the loose sand and in the dense sand.
Looking at the volumetric strain results (Davidson and Boghrat, 1983), the loose
sand undergoes expansion (negative volumetric strain) below the tip and immediately
adjacent to the side of the cone, whereas contraction of soil is observed only farther from
the side. This is also true for the dense sand except the occurrence of stronger
densification at points adjacent to the probe. The maximum displacements recorded were
74
4.0 mm in the loose sand and 5.0 mm in the dense sand. The zones of influence were of
radius 125 mm and 155 mm horizontally and 125 mm and 100 mm below the tip
vertically, in the loose sand and in the dense sand, respectively.
A more recent study of sand by Hughes and Robertson in 1985 constituted a
complementary study to the deformation analysis by Davidson and Boghrat (1983).
Their work was focussed on state of stress around the full displacement pressuremeter
probe. This type of pressuremeter probe has at its tip the standard cone penetrometer
CPT (Hughes and Robertson, 1985). Results of the qualitative analysis from the
McDonald’s Farm data showed that a very high stress is developed where the tip passes,
but then the stress drops significantly when the zone is in contact to the cone sleeve.
Figure 3-5. Stresses around the cone penetrometer. (Reprinted by permission of CanadianGeotechnical Journal, from Hughes, J. M. O., Robertson, P. K., FullDisplacement Pressuremeter Testing in Sand, Canadian GeotechnicalJournal, Vol. 22, No. 3, pp. 298-307, 1985)
75
The phenomenon is illustrated in Figure 3-5, where the tip resistance reaches
6500 kPa whereas the estimated average lateral stress on the cone sleeve is only 90
kPa.Hughes and Robertson (1985) also analyzed the stress paths during the process of
penetration. The change in insitu state of stress can be seen clearly using the reference
points of Figure 3-6A. Starting from point A, the soil quickly reaches a failure state, then
as the cone penetrates more, an unloading of the adjacent points C begins and the state of
stress stabilizes at point E. The unloading point E corresponds to the negative volumetric
strain observed in the deformation analysis by Davidson and Boghrat (1983).
A BFigure 3-6. Stress paths and distribution around cone penetrometer. A) Stress path. B)
Stress distribution. (Reprinted by permission of Canadian GeotechnicalJournal, from Hughes, J. M. O., Robertson, P. K., Full DisplacementPressuremeter Testing in Sand, Canadian Geotechnical Journal, Vol. 22, No.3, pp. 298-307, 1985)
76
In terms of magnitude, Figure 3-6B indicates the zone of high residual stress
where the radial stress and the circumferential stress are very high; marked by the point
E, those points correspond to the zone of high densification of around 6% volumetric
strain in Davidson and Boghrat (1983).
Indeed, the insertion of the cone penetrometer in the ground causes soil
disturbance, and hence change of the initial state of stress. Theoretical studies of the
cavity expansion have been done by researchers over the years and they are briefly
discussed in the section on the Pressuremeter test.
Correlating Soil Properties from CPT Data
The CPT has a good repeatability of data because of the more uniform standard
procedure and more uniform equipment, as opposed to the SPT. However, the estimation
of engineering properties still needs improvement especially for the calibration of the
equipment, and the disturbance effects due to the cavity expansion during the insertion.
Lunne et al. (1997) gave information on the applicability and the reliability of the CPT to
estimate the different properties of the soil. It was then realized that generally the CPT is
more reliable for estimating the shear strength parameters whereas it is less reliable for
the insitu state of stress parameters and the stiffness parameters.
Soil classification
Soil type behavior. One of the advantages of using the CPT is the soil
classification and stratigraphy of the soil profile. The soil identification usually relates
the tip resistance with the friction ratio (%). Schmertmann (1978) created a correlation
from using the mechanical CPT; others are Douglas and Olsen (1981), and Roberston et
al. (1986). The soil identification correlations refer more to the classification of the soil
behavior type rather than grain size and mineral composition of the soil.
77
Relationship between CPT qc and SPT N. The several correlations that have
been established from SPT and CPT enabled few authors to relate the SPT N blow count
with the CPT tip resistance qc. The relationship is linked with the grain size distribution
of soil as Robertson et al. (1986a) found out. Their correlation was updated by other
investigators such as, Jamiolkowski et al. (1985). Other forms of the correlation are also
available in Lunne et al. (1997), and Kulhawy and Mayne (1990). As an example for
such relationship, Robertson et al. (1983) made a table based on their soil.
Correlations for cohesionless soils
Physical properties. A rough approximation of the unit weight is estimated
based on the soil classification by Robertson et al. (1983). To each soil zone is given a
value of unit weight. As for the relative density Dr, several authors have proposed charts
and equations. In many cases, the effective horizontal stress 0hσ or the effective vertical
stress 0vσ are also included in the correlations besides the tip resistance qc. The chart
proposed by Robertson and Campanella (1983) accounts for the location of the sand.
Another type of correlation is a chart by Jamiolkovski (1985), which was
transformed into Eq. (3.29). This correlation includes sands from Ticino, Ottawa, Edgar,
Hokksund and Hilton; and resulted from calibration chamber tests (Kulhawy and Mayne,
1990).
−
×= 1log68
0va
cr
pq
Dσ
(3.29)
The calibration chamber tests were extended to overconsolidated sands and new
correlations were derived by Kulhawy and Mayne (1990). The influences of the
78
overconsolidation ratio OCR and the normalized tip resistance are represented in Eq.
(3.30) given below
5.00
18.02
3051
××
=
a
v
a
c
Cr
p
pq
OCRQD
σ(3.30)
where QC is a compressibility factor such that, QC = 0.91 for high compressibility, QC =
1.00 for medium compressibility, QC = 1.09 for low compressibility. In the practice, QC
is sometimes taken to 1.0 (e.g., Hortvah, 1994).
Insitu state of stress. Direct relationships between the tip resistance qc and the
insitu state of stress parameters have not been established for cohesionless soils. This is
due mainly to the diffult sampling of sand materials. The overconsolidation ratio OCR or
the coefficient of lateral stress K0 is usually estimated via the relative density of the soil,
Eq. (3.30) or the effective friction angle (Kulhawy and Mayne, 1990; Marchetti, 1985).
Kulhawy et al. (1989) obtained from calibration chamber tests the correlation, Eq. (3.31),
that relates the effective horizontal stress, 0hσ , to the tip resistance qc. It is indirectly a
correlation for the coefficient of lateral stress, K0.
=
20exp35
25.1
0
r
a
c
a
h
Dpq
pσ (3.31)
When combined with Eq. (3.30) to estimate K0, Eq. (3.31) gave a good match for the case
of Stockholm sand, compared to the K0 from SPT and PMT (Kulhawy et al., 1989). A
useful relationship between the coefficient of lateral stress K0 and the overconsolidation
79
ratio OCR was provided by Mayne and Kulhawy (1982). The laboratory-based
relationship, Eq. (3.32) is applicable for both sands and clays
( ) tcsintc0 sin1 φφ OCRK ×−= (3.32)
and also applies for soils which have experienced simple loading-unloading history.
Strength parameters. Unlike the insitu state of stress, the friction angle for
sands has been correlated with the tip resistance qc by several authors. The most
common is the chart by Robertson and Campanella (1983) for normally consolidated and
uncemented sand. This correlation tends to overestimate the value of the friction angle
φtc. Usually, the correlation takes into account the effect of the effective overburden
pressure 0vσ .
Other correlations are from Debeer (1974), Janbu and Senneset (1974),
Durgunoglu and Mitchell (1975), Schmertmann (1978), etc. Based on the author’s
experience, the different methods estimate the friction angle within error of 5%, which is
a good sign of consistency. A more accurate correlation is presented in Eq. (3.33). It
was obtained from calibration chamber test and is for both normally consolidated and
overconsolidated sands (Kulhawy and Mayne, 1990).
)8.2.. ;64.0 ;633( log0.116.17 25.0
0
tc °===
+= DSrn
p
pq
a
v
a
c
σφ (3.33)
Kulhawy and Mayne (1990) mentioned that this correlation is consistent with those for
relative density in Eq. (3.30) and coefficient of lateral stress at rest from Eq. (3.31).
Stiffness parameters. Both the tangent constrained modulus Mt and the secant
Young’s Modulus E50 have been estimated from the CPT basic parameters. Evaluation of
80
the stiffness parameters from the CPT cannot be fully reliable because of the following
three reasons (Lunne et al., 1997): 1) modulus depends on effective stress and stress
history, 2) the insitu test conditions, stress level, drainage and direction of loading cannot
be controlled, and 3) reference modulus values are rarely or seldom documented. A chart
by Bellotti et al. (1989) to estimate the secant modulus Es, is considered reasonable
according to Lunne et al. (1997).
Other correlations that take into account the stress level were developed by
Robertson and Campanella (1983) for secant Young’s modulus E50, and for the
constrained modulus Mt. For practical use, linear relationships are more common as a
first order estimation. Among others, Lunne and Christophersen (1983) proposed:
for normally consolidated sands
MPa50for MPa120MPa5010MPafor [MPa] 202
MPa10for 4
0
0
0
>=<<+=
<=
c
cc
cc
qMqqM
qqM(3.34)
and for overconsolidated sands
MPa50for MPa250MPa50for 5
0
0
>=<=
c
cc
qMqqM
(3.35)
Bowles (1996) has given the linear type of relationship for normally consolidated sand
cqE 4) to(2 s = (3.36)
For overconsolidated sands, the author would use Eq. (3.19) mentioned in SPT section.
In dynamic problems, Eq. (3.37) by Imai and Tonouchi (1982) can be used to
estimate the dynamic shear modulus of the soil Gmax.
[bars]in 5.4
125661.0
max cc q
qG
= (3.37)
81
Mayne and Rix (1995) also proposed a correlation from a seismic cone or a piezocone
between the shear modulus and the qc corrected for pore pressure qT (Bowles, 1996).
Correlations for cohesive soils
Physical properties. The consistency of cohesive soils related to the tip
resistance qc was given by Szechy and Varga (1978) in a table. From their estimation, a
clay soil having a tip resistance qc less than 5bars is considerd to be a very soft clay (CI <
0.5); on the contrary, tip resistance qc over 60bars is a very hard clay with consistency
index greater than 1.5.
Insitu state of stress. Clays are good materials for undisturbed samples. The
CPT appears suitable for estimating the insitu state of stress of clays. The correlations
have better statistical characteristics than those from SPT (Kulhawy and Mayne, 1990).
The first correlation, Eq. (3.38) relates the preconsolidation stress cσ with the tip
resistance qc as follows (Mayne, 1986).
)31.2.. ;858.0 ;113( 29.0 2acc pDSrnq ====σ (3.38)
From using the piezocone, an even better correlation is obtained; it includes the
correction on the tip resistance for 11 Canadian clays (Mayne and Holtz, 1988)
( ) )02.1.. ;904.0 ;74( 33.0 20 avTc pDSrnq ===−= σσ (3.39)
In addition to the preconsolidation stress, correlations with the overconsolidation ratio,
OCR are also common in clays; for instance, Mayne (1990) obtained from 52 clays the
following
( ))76.0.. ;762.0 ;161( 32.0 2
0
0 ===−
= DSrnq
OCRv
vT
σσ (3.40)
82
For the determination of the coefficient of lateral stress, K0, it is worth remembering that
the K0-OCR relationship in Eq. (3.32) applies to cohesive soils as well.
Strength parameters. The undrained cohesion su is derived from the theoretical
equation of bearing capacity, Eq. (3.41), referring to the tip resistance qc
0 vukc sNq σ+×= (3.41)
where σv0 is the total overburden stress, and Nk the cone factor which ranges as follows:
307 << kN (3.42)
this range includes the various existing cases, for instance Senneset et al. (1988)
proposed 15 to 30 for CAUC1; 7 to 13 for cavity expansion theory, and 14 to 18 for
steady penetration theory (Kulhawy and Mayne, 1990).
Stiffness parameters. Few studies were achieved to correlate the CPT qc with
the elastic properties of clays. It is more common to first determine the value of the
undrained cohesion su and then use the different relations such as Eq. (3.26) to obtain the
elastic modulus. Sanglerat (1979) provided a table containing the proportion αM
(depending on the plasticity of the clay) between the constrained modulus M and the cone
tip resistance qc (mechanical cone only) (UF Course Notes, 2000). Bowles (1996)
proposed for very soft clays and silts the following correlation
cqE 8) to(3 s = (3.43)
Advantages and Disadvantages of CPT
The main advantange of CPT is the repeatability of the data. The test procedure is
more uniform and so are the testing equipment. Other advantages are that the CPT is
faster, less expensive and is becoming more widely used. Its reliability of use in soft
1 Anisotropic Consolidated Undrained Triaxial Test Sheared in Compression.
83
soils, and ability to identify soil type in a sounding profile are of significant advantage.
The errors from different factors such as, operator is greatly reduced and the continous
data recording is very convenient for further interpretation of the sounding. One of the
disadvantages of the CPT is that no samples can be obtained. The CPT has also limited
capability to penetrate hard soils such as, cobbles, boulders, gravels, and the like.
Flat Dilatometer Test: DMT
Background and Procedure
The Flat Dilatometer test, also called the Marchetti Dilatometer test, is the most
recent insitu test used in this study. Dr. Silvano Marchetti, a professor at L’Aquila
University, Italy, developed the Flat Dilatometer in 1974 (Marchetti, 1975). Originally,
the Flat DMT was a device to measure the insitu soil modulus for laterally loaded steel
piles. But the extensive field testing produced a number of correlations which made the
Flat DMT one of the major geotechnical insitu tests to determine the various soil
properties. The Flat DMT was introduced in the U.S.A in 1980 and grew rapidly with the
major contributions by researchers such as Marchetti (1980), Schmertmann (1986), and
Bullock (1983). In addition to the determination of the soil properties, the DMT
settlement calculation of shallow footings in sand is widely used in geotechnical
engineering. The DMT equipment consists of three major parts: the control unit, Figure
3-7, the probe, Figure 3-8, and the pneumatic-electrical connection cable.
As shown in Figure 3-8, the probe is a 96 mm wide, 15 mm thick stainless steel
blade with a thin, flat, circular, expandable steel membrane on one side. The test is
performed by pushing the probe through push rods (with friction reducer at the tip) to the
desired depth; then by inflating the expandable thin membrane by manual operation on
the control unit.
84
Figure 3-7. DMT control unit (dual gage type). (Source: Schmertmann, 1988)
Figure 3-8. DMT probe. (Source: Marchetti, 1999)
The DMT has not been standardized by the ASTM standard yet, despite the
suggested method submitted by Schmertmann in1986. The most common outline of the
procedure is given below. The DMT always starts with the calibration of the membrane.
The calibration is necessary in order to obtain the correction parameters ∆A and ∆B.
85
Initial position of the membrane is stress free, thus ∆A is the external pressure required to
make the membrane flush with the body of the blade (A-position), pulling the syringe in
the configuration of will allow to reach the A-position ∆A; on the other hand, ∆B is the
internal pressure that lifts the membrane to a distance of 1.10 mmfrom the stress free
position (B-position), pushing the syringe to reach B-position yields to reading ∆B. More
details on the calibration procedure and acceptable values of ∆A and ∆B are easily
available in the literature (e.g., Marchetti, 1999; Schmertmann, 1988).
After the calibration, preliminary tasks are of great importance as well: making
sure there is no leaks in the pneumatic cables and the different connections (Nitrogen
tank-control unit, control unit-probe) after all necessary threading and connections have
been made; making sure the electric current flows during the test. This is accomplished
by connecting the ground plug-in with the blade and running, for instance, ∆A again, the
audio buzzer, Figure 3-7 on is the sign of existing current circuit. The actual DMT is
carried out normally every 200mm of depth. Excluding the operation details, the four
major steps are, Figure 3-9: 1) pushing or driving the probe to the desired depth, 2) taking
the A reading, 3) taking the B reading, and 4) taking the C reading (optional). After
pushing the probe to the desired depth, the membrane is pushed in by the soil (similar to
A-position) and the audio buzzer is on, the A reading is then obtained by inflating the
membrane to be lifted off from the A-position and the buzzer is off; by keeping
pressurizing the membrane, the B-position is reached (membrane movement 1.10 mm)
and the buzzer turned off again, this is the B reading; after the B reading is taken,
86
INFLATE DEFLATE
1 2 3 4
PUSH or DRIVE
AUDIO BUZZER →
READING →ON OFF ON (OFF) ON
P A B C
1.10mm
Figure 3-9. Steps during procedure of DMT. (Source: Schmertmann, 1988)
the system is vented quickly but then more slowly to take the C reading when the buzzer
is back on, that is the A-position is reached. The working principle in the audio buzzer-
related-membrane deformation is illustrated in the work by Marchetti et al. (2001).
The equipment used to insert the blade in the ground is the drill rig or the cone
truck. The University of Florida uses the cone truck in Figure 3-4 with the rate of
penetration 20mm/s. Other penetration rates are accepted too for the DMT—e.g. the
Eurocode (Marchetti et al., 2001).
Disturbance Due to Insertion of Blade
In addition to the axisymmetric study on the disturbance effect due to insertion of
the cone penetrometer in CPT, researchers have attempted to describe qualitatively and
quantitatively the effect of the insertion of the DMT blade into the ground. The study is
neither axisymmetric nor plane strain, but can be assumed as a plane strain type of
problem. The wedge shape and lower 16° apex angle at the DMT blade should induce
87
much less penetration disturbance than the 60° axisymmetric CPT tip. Other phsyical
dimensions are the diameter 35.7 mm of the CPT cone versus the flat plate of
235 mm×96 mm×15 mm of the DMT. Baligh and Scott (1975) carried out theoretical
and experimental comparative studies on the penetration of different apex angles in clays.
In their work, Baligh and Scott (1975) show the plane strain deformation patterns in the
initially square grids. They obtained the following major conclusions: the soil
displacement is less for a smaller apex angle than it is for a larger apex angle; for the
same width of the wedges, larger wedge angle poduces higher straining levels and
smaller plastic zone; the mechanism of the penetration in sharper wedges can be more
adequately represented by the plasticity theory.
The tests that Davidson and Boghrat (1983) performed in the loose and dense
sands, using the cone penetrometer (axisymmetrical) and the DMT blade (plane strain),
yielded the same conclusions. The direction of the displacements are the same in the
loose sand and in the dense sand for both probes. However the magnitudes are smaller
and more uniform for DMT blade than for the cone penetrometer. For instance, 2.2 mm
versus 4.0 mm in the loose sand and 3.6 mm versus 5.0 mm in the dense sand. The
volumetric strains are also smaller and more uniform around the DMT blade than those
around the CPT cone. Zones of loosening (negative volumetric strain) are common
below the tip for both probes, either in the loose or the dense sand, but these zones are
also present at points adjacent to the cone. Finally, in both sands, the zone of influence is
larger for the cone penetrometer than for the DMT blade. For instance, measurable CPT
displacements approach a horizontal distance of 125 mm (loose sand) and 155 mm (dense
sand) versus 100 mm (loose sand) and 160 mm (dense sand) for the DMT.
88
Marchetti (1981) has proposed, Figure 3-10, the stress-strain curve during the
penetration and the membrane expansion; it is similar to the unload-reload modulus of
the Pressuremeter test, that is, the soil becomes stiffer for a smaller displacement of the
soil, portion CD.
A
B
C, D
B
A
C
D
Stre
ss
Strain
A B C PenetrationC D Membrane expansion
Figure 3-10. Proposed stress-strain curve during DMT. (Source: Marchetti and Crapps,1981)
Based on these analyses, the insertion of the DMT blade into the ground causes
less soil disturbance, but not negligible, than the CPT cone penetrometer. The difference
is attributed primarily to the two parameters: the angle apex: 16° for the blade and 60° for
the cone, and the thickenss of the blade 14.0 mm versus the radius of the cone 35.7 mm.
Correlating Soil Properties with DMT Data
A large advantage of choosing the DMT, over any other insitu tests is the
uniformity of the DMT equipment, uniformity of procedure, and uniformity of
correlations. Local correlations are sometimes developed for more accurate DMT
89
estimates of soil properties, but for most soils, the original Marchetti correlations are
sufficiently accurate (Schmertmann, 1986). The basic DMT parameters are: the
Dilatometer Modulus ED, the Horizontal Stress Index KD, and the Material Index ID. The
Pore Pressure Stress Index UD (seldom used). They are obtained from the two pressure
values: p0 and p1, which are calculated from the corrected values of the data readings A
and B, as presented in the upper part of Table 3-2. The material index ID is an indication
of the soil type according to Marchetti (1980). The relations, given in Eq. (3.44) provide
a useful description of the soil behavior. Marchetti et al. (2001) noted that misdescription
between clay and silt sometimes occurred in identifying the soil with ID.
)10(.81 sand8.16.0 silt 6.01.0 clay
<<<<<<
D
D
D
III
(3.44)
The horizontal stress index KD is a stress history parameter and therefore a good
estimation of the OCR profile from the DMT sounding. The dilatometer modulus ED is a
parameter indicative of the soil compressibility. However, Marchetti et al. (2001) found
that ED has to be used in combination with ID and KD, and should not be considered as
closer to the ordinary elastic modulus of the soil. Many soil properties are estimated
from KD and ED. Some of the soil engineering parameters are presented in the lower part
of Table 3-2. Other correlations are presented in the following sections.
Soil classification and physical properties
Soil type diagram. Marchetti and Crapps (1981) have established the chart that
provides the soil type and unit weight from the Dilatometer modulus ED and the material
index ID, Figure 3-11.
90
Table 3-2. Dilatometer parameters from the raw data A and B
SYMBOL DESCRIPTION BASIC DMT REDUCTION FORMULAEp0 Corrected First Reading p0 = 1.05 (A - ZM + ∆A) - 0.05 (B - ZM - ∆B)
p1 Corrected Second Reading p1 = B - ZM - ∆BZM = Gage reading when vented to atm.If ∆A & ∆B are measured with the samegage used for current readings A & B,set ZM = 0 (ZM is compensated)
ID Material Index ID = (p1 - p0) / (p0 - u0) u0 = pre-insertion pore pressure
KD Horizontal Stress Index KD = (p0 - u0) / σ'v0 σ'v0 = pre-insertion overburden stress
ED Dilatometer Modulus ED = 34.7 (p1 - p0) ED is NOT a Young's modulus E. EDshould be used only AFTER combining itwith KD (Stress History). First obtainMDMT = RM ED, then e.g. E ≈ 0.8 MDMT
K0 Coeff. Earth Pressure in Situ K0,DMT = (KD / 1.5)0.47 - 0.6 for ID < 1.2
OCR Overconsolidation Ratio OCRDMT = (0.5 KD)1.56 for ID < 1.2
cu Undrained Shear Strength cu,DMT = 0.22 σ'v0 (0.5 KD)1.25 for ID < 1.2
Φ Friction Angle Φsafe,DMT = 28° + 14.6° log KD - 2.1° log2 KD for ID > 1.8
ch Coefficient of Consolidation ch,DMTA ≈ 7 cm2 / tflex tflex from A-log t DMT-A decay curve
kh Coefficient of Permeability kh = ch γw / Mh (Mh ≈ K0 MDMT)
γ Unit Weight and Description (see chart in Figure 16)
MDMT = RM ED
if ID ≤ 0.6 RM = 0.14 + 2.36 log KD
if ID ≥ 3 RM = 0.5 + 2 log KD
if 0.6 < ID < 3 RM = RM,0 + (2.5 - RM,0) log KDwith RM,0 = 0.14 + 0.15 (ID - 0.6)
if KD > 10 RM = 0.32 + 2.18 log KD
M Vertical Drained ConstrainedModulus
if RM < 0.85 set RM = 0.85
u0 Equilibrium Pore Pressure u0 = p2 = C - ZM + ∆A In free-draining soils
91
I )EQUATION OF THE LINES:
SOIL DESCRIPTION
0.6
Material Index
If PI>50, reduce by 0.1
D
Dila
tom
eter
Mod
u lus
0.1
and/orPEAT
5
MUD1210
20
50
( )1.5
0.2 0.5
MUD
A
B
C
0.33
1.6
1.8
1.7
1000
(bar
)E
100
200
D 500
2000
D
0.585
0.6570.694
CLAY
2.05
DC
AB 0.621
m
1.9
SILTY
2.0132.2892.564
1.737n
E =10(n+m log
3.3
1.7
γ
1
I2
D
0.8 1.2
1.6
1.7
1.8
5
SAND
2
1.8
1.9
2.15
1.95
1.8
2.1
SILT
CLA
YEY
SILTY
SAN
DY
D
and ESTIMATED γ/γw
Figure 3-11. Soil type and unit weight from DMT. (Source: Marchetti and Crapps, 1981)
The above chart has been updated by other authors (e.g., Lacasse and Lunne, 1988;
Marchetti, 1988); however it is still considered as a good estimation of the soil profile.
A correlation between the horizontal stress index KD and the relative density Dr
for normally consolidated and uncemented sands was also found in Kulhawy and Mayne
(1990). It is from only four datapoints from chamber tests and two datapoints from field
tests. Obviously, more data are needed to reduce the uncertainty of such correlations.
Relationship between DMT data and SPT N. Schmertmann (1988) provided a
rough approximation of the relation between the dilatometer constrained modulus M and
the SPT N, based on data from Gainesville sites. Schmertmann pointed out that the
relationship cannot be reliable in that they are site specific and equipment specific. For
indication purpose, Eq. (3.45) is presented
SPTDMT NM 30[bars] = (3.45)
92
Correlations for cohesionless soils
Insitu state of stress. For the estimation of the overconsolidation ratio OCR,
Bullock (1983) reported Schmertmann’s modification of the Kulhawy and Mayne (1982)
equation previously presented in Eq. (3.32), in order to fit results from laboratory test
data, Eq. (3.46).
−
=tcsin8.0
1
tc
0
sin1
φ
φK
OCR (3.46)
In the estimation of the coefficient of lateral stress K0 in sands, the combination
with the CPT tip resistance qc is a common method. It has been produced by authors
such as, Baldi et al. (1986) from calibration chamber tests, or Marchetti (1985) from field
tests:
00 0046.0095.0376.0
v
cD
qKK
σ−+= (3.47)
00 00093.0071.0359.0
v
cD
qKK
σ−+= (3.48)
respectively. Marchetti et al. (2001) suggested that despite the good agreement with the
Self-boring Pressuremeter data in Eq. (3.47), caution should be exercised when dealing
with overconsolidated or cemented sands. Schmertmann (1988) also derived a
correlation from chamber test data obtained for both normally consolidated and
overconsolidated sands using Jaky’s equation (normally consolidated) to derive Eq.
(3.50). The insitu parameter K0 is a function of the horizontal stress index KD and the
axisymmetric friction angle φtc.
Equation) s(Jaky' sin1 tcNC0 φ−=K (3.49)
93
( ) ( ) ( )( )tc
2tctctc
0 sin1717192sin1717sin1152sin1232340
φφφφ
−−−−−+−×−+
= DD KKK (3.50)
the friction angle φtc being obtained from Eq. (3.51) below. Equation (3.50) is considered
more reasonable as it includes the effect of frictional strength (Schmertmann, 1988).
Strength parameters. Three methods are currently used to calculate the friction
angle φ. The first one is given in Table 3-2, it is considered to be a conservative method
for evaluation of φ; the second one is an iterative procedure by Schmertmann (1982)
given in Eq. (3.51)
( ) Hf FZSRODWTqDFRICBDFRICDMAREA
uRODDIAMTHRUST
+×+×
×−+
−××−=
24
019.142
tan
2
02ps
π
πφ
(3.51)
where the friction angle is dependent on the dimensions of the push rods and the blade,
the thrust force, and the lateral forces acting of the blade. Details of this rational method
by Schmertmann and the variable names can be found in the FHWA Report by
Schmertmann (1988). One should notice the use of the drained plane strain friction angle
φps in lieu of the axisymmetric friction angle φtc in Eq. (3.51). The relation given in Eq.
(3.14) is used to obtain the desired friction angle; the third one is a graphical solution of
the theory by Durgunoglu and Mitchell (1975) and using both the cone tip resistance qc
and the coefficient of lateral stress K0.
Stiffness parameters. The main DMT elastic parameter is the drained, one-
dimensional (Oedometer type) tangent modulus at 0vσ , M. The constrained modulus M
is proportional to the dilatometer modulus ED by the factor RM in Table 3-2. The method
for detemination of the constrained modulus M is provided for different values of ID and
94
KD. Subsequently, the drained modulus Et can be computed from the constrained
modulus M.
( ) ( )( ) ME ×
−−×+
=ν
νν1
211 (3.52)
Moreover, the dilatometer modulus ED is related to the drained Young’s modulus by
( ) EED ×−
=21
1ν
(3.53)
Correlations for cohesive soils
Insitu state of stress. The preconsolidation stress cσ was found by Mayne
(1987) to be related with the DMT initial contact stress p0.
( ) )15.1.. ;896.0 ;76( 51.0 200 ac pDSrnup ===−=σ (3.54)
For the overconsolidation ratio OCR, few correlations have been developed.
Marchetti (1980) provided the original equation in Table 3-2 for uncemented cohesive
soils. The same equation was slightly modified by authors such as, Lacasse and Lune
(1988) into
1.67 1.35 where 225.0 <<= mKOCR mD (3.55)
in which lower values of m are used for high plasticity clays and vice versa. Another
correlation of the same form was made by Kamei and Iwasaki (1995).
43.134.0 DKOCR = (3.56)
The coefficient of lateral stress K0, is obtained similarly. The original equation in
Table 3-2 was developed by Marchetti (1980), which then was modified by Lacasse and
Lunne (1988) (Marchetti et al., 2001) into
0.64 0.44 where 34.00 <<= mKK mD (3.57)
95
in which lower values of m are used for high plasticity clays and vice versa. Also,
estimation of K0 by using Eq. (3.32) by Kulhawy and Mayne (1982) is valid for cohesive
materials.
Strength parameters. The undrained shear strength, su, is stress level dependent
and is a function of the lateral stress index, KD, as Eq. (3.58), by Marchetti (1980), shows
( ) 25.10 5.022.0 Dvu Ks ×= σ (3.58)
Marchetti et al. (2001) discusses that Eq. (3.58) includes with a good accuracy the
different data points from various researchers such as Lacasse and Lunne (1988), and
Powell and Uglow (1988).
Stiffness parameters. In cohesive soils, the elastic parameter from the DMT is
the tangent constrained modulus M. Table 3-2 includes the determination of the
constrained modulus M for cohesive soils. To convert from the tangent constrained
modulus M to the undrained Young’s modulus Eu
( ) ( )( ) ME u
u ×−
−×+=
ννν
1211 (3.59)
where νu = 0.5 is the undrained Poisson’s ratio.
Advantages and Disadvantages of DMT
The DMT causes less disturbance of the soil than the CPT does (Baldi and Scott,
1975; Davidson and Boghrat, 1983) and is one of the only tests to directly measure the
lateral stress (Others include the Self-boring PMT, the Handy K0 Blade and Hydraulic
Fracture tests). The DMT is simple and rapid, and the equipment can be used in a wide
variety of soils. Rapid data reduction is possible as well using the currently available
programs—e.g., Dilatometer (GPE, Inc.). It is slowly gaining acceptance among
engineers as the correlation database is expanded. Experience with the equipment is
96
limited for cases of very dense or cemented soils (Kulhawy and Mayne, 1990) and
damage to the blade is likely to occur during penetration into these soils.
Pressuremeter Test: PMT
Historical Backgound and Equipment
By the definition given by Clarke (1995), the pressuremeter “is a cylindrical
probe that has an expandable flexible membrane designed to apply a uniform hydraulic
pressure to the walls of a borehole”. Therefore, the pressuremeter test is the
measurement of the soil pressure-deformation response resulting from the expansion of
the probe. The very first cylindrical expandable device made was monocell, and was
officially introduced by Kögler in 1933, Figure 3-12. Kögler’s work on the
pressuremeter did not have so much impact to the engineering world although he
produced pressure-deformation curves in fine-grained clays and sands. About two
decades later, in 1954, Ménard invented his prototype pressuremeter in France, and then
carried out his first test at the University of Illinois under the direction of Professor Ralph
Peck. The Ménard pressuremeter greatly differed from Kögler’s because it has three cells
stacked in the cylindrical probe. Two guard cells (top and bottom) are to protect the
middle cell, which is the pressuremeter test actual cell, from end effects caused by the
finite length of the probe. Thus, the idea of plane strain (zero vertical deformation) was
applied as there is only radial expansion of the cavity, and the height of the cylinder was
assumed to be infinite (Baguelin et al., 1978).
97
Figure 3-12. Sketches of Kögler and first Ménard probes. (Reprinted by permission ofTrans Tech Publications, from Baguelin, F., Jezequel, J. F., Shields, D. H.,The Pressuremeter and Foundation Engineering, Trans Tech Publications,1978)
As further development of the Ménard PMT, Ménard tried to calibrate the PMT
data with the more common Mohr-Coulomb soil parameters c and φ, with marginal
success. The basic parameters was the limit pressure pl. In the mid-1960s, the
Laboratoire des Ponts et Chaussées designed a new method of installing the probe to
minimize borehole disturbance and improve test results: the Self-boring Pressurementer.
The Self-boring PMT was developed in France and United Kingdom, and first
used in 1967. The resulting pressure-deformation curves indeed differed from those from
the Ménard PMT, difference in shape of the curves, pressuremeter modulus, and in limit
pressure. In 1975, the British Governement, Department of Energy financed the
development of the Push-in pressuremeter. These probes were intended for use in caving
(saturated sands), and soils difficult (hard) to drill with the Self-boring PMT. The Push-
98
in PMT or full displacement PMT was initially used for offshore investigations, but it has
grown rapidly since its commercial launch in 1980, and expanded to onshore
investigations. Various types of Push-in PMT are now in use. The cone PMT is a
monocell probe mounted behind a standard cone penetrometer (Fugro type). The Pencel
PMT is a monocell probe with a cone at the tip (see Figure 3-13), which is also pushed
into the ground using the cone truck. The Push-in PMT causes full displacement of the
soil. In contrast to the Ménard PMT, where stress relief occurs during borehole
preparation, the cone tip overconsolidates the surrounding soil during penetration and
then allows relaxation.
Figure 3-13. Push-in (Pencel) Pressuremeter probe. (Reprinted by permission of ASTM,from Briaud, J.-L., Shields, D. H., A Special Pressuremeter andPressuremeter Test for Pavement Evaluation and Design, GeotechnicalTesting Journal, ASTM, Vol. 2, No. 3, pp. 143-151, 1979)
At the present time, there are three categories of pressuremeter, each of the
categories has more than two types of probe: the Pre-bored PMT (Ménard types), the
99
Self-boring PMT, and the Push-in PMT. The resulting pressure-deformation curves of
the soils are different.
The three categories of pressuremeters have different control units connected to
their probes during the tests. The Self-boring PMT has the most sophisticated and
complicated components. The Ménard PMT alone has different versions of control unit
(compatible with different types of probe: e.g., GC, GB, and GA probe), whereas the
Pencel PMT has the simplest form of the control unit. Details of the equipment can also
be found in the literature (e.g., Baguelin et al., 1978; Clarke, 1995). The following
section is a brief introduction of the cavity expansion theory. The theory is closest to the
Self-boring PMT principle. The disturbance effect due to insertion of the cone in Push-in
PMT is already discussed in the CPT section.
Cavity Expansion by Pressuremeter Probe
Various researchers have dealt with the theory of cavity expansion since the 19th
century. The oldest theoretical analysis was provided by Lamé in 1875 (Clarke, 1995),
assuming a linear elastic material around the hollow cylinder. Ménard in 1957, and more
recently Papanastasiou and Durban in 1997, chose a more soil mechanics oriented
approach using linear elastic perfectly plastic behavior. Other authors, such as Denby in
1978, and Ferreira and Robertson in 1992, went further by considering nonlinear elastic
perfectly plastic. The hypothesis of a hardening soil constitutive model has not been
analyzed so far. The basic idea is that the soil is divided into a plastic region, and an
elastic region, and elastic region when expanded by the uniform pressure from the
pressuremeter probe. The length of the cylindrical probe is also assumed infinite so that
the plane strain condition applies in the analysis (vertical strain εz = 0). In the work by
100
Houlsby et al. (1995), the state of stress around the probe is shown with respect to the
failure line. The radial stress σr and the hoop stress σθ have initially the same magnitude
and the medium is elastic. But when the pressure is applied, plastic regions start to form
and eventually reach the cylindrical region of radius r1. Subsequent unloading will cause
some points in the plastic region to regain the elastic state while others remain plastic.
The analysis by Houlsby et al. (1995) is for sands.
Figure 3-14. Stress paths and stress-strain curves in PMT. (Reprinted by permission ofCanadian Geotechnical Journal, from Hughes, J. M. O., Robertson, P. K.,Full Displacement Pressuremeter Testing in Sand, Canadian GeotechnicalJournal, Vol. 22, No. 3, pp. 298-307, 1985)
101
Hughes and Robertson (1989) provided a comprehensive description of the stress
paths followed by a point adjacent to the probe membrane in sands, and the pressure
expansion as a function of the probe installation method. Figure 3-14 shows that the
stress relaxation due to the prebored hole reduces the radial stress σr more than the hoop
stress σθ. On the contrary, for the Push-in PMT, the insertion of the probe produces more
changes in the hoop stress than in the radial stress. For both cases, the initial state of
stress has been changed by the method of installation, and this is reflected in the pressure-
deformation curves on the right hand of Figure 3-14.
A
B
Figure 3-15. Stress distribution in linear elastic-perfectly plastic soils. A) Purely cohesivesoil. B) Soil with friction and cohesion. (Reprinted by permission of TransTech Publications, from Baguelin, F., Jezequel, J. F., Shields, D. H., ThePressuremeter and Foundation Engineering, Trans Tech Publications, 1978)
102
Quantitative analysis by Ménard in 1957, for elastic-perfectly plastic soil is
presented in Figure 3-15. The spatial variation of radial stress σr and the circumferential
stress σθ are given at the time when the limit pressure pL is reached. The pressure p0
indicates the initial state of stress. The variation of the stresses is hyperbolic for both
sands and clays in the elastic and plastic zones, except for clays in the plastic zone where
a spiral logarithmic variation is assumed.
Pressuremeter Test Procedures
The PMT has been standardized in the U.S.A by ASTM D 4719 since 1987. The
standard equipment, calibration procedures, testing procedures and other specifications
(for example, the allowable size of the borehole with respect to the probe diameter, etc.)
are provided in ASTM D 4719. The procedures for Ménard PMT and the Self-boring
PMT are also widely available in the pressuremeter literature (Baguelin et al., 1978;
Clarke, 1995). The Self-boring pressuremeter is the most sophisticated geotechnical
insitu device. For the Cambridge Self-boring PMT, a paper by Professor John Davidson
(1983) at the University of Florida is suggested for a complete equipment description and
detailed test procedure. The emphasis herein will be on the Push-in PMT because of its
use throughout this research.
More than the equipment operation, the test procedure and data reductions are
very critical during the Push-in Pressuremeter test. The Ménard PMT is stress controlled
whereas the Self-boring PMT can be performed either stress-controlled or strain-
controlled. For the Push-in PMT, generally, the loading portions of the PMT pressure-
deformation curves are performed strain-controlled and the most common rate of strain
used is between 1.0%/min and 4.0%/min (Clarke, 1995; Powell and Shields, 1995). The
103
upper boundary corresponds to the strain rate of 1 revolution per 9 seconds when
cranking the Pencel PMT control unit to expand the PMT probe. The unload-reload parts
are stress-controlled and are performed at a rate of strain smaller than that during the
loading; for instance, strain rate was decreased to 0.03%/min during the PMT in sands
reported by Nutt and Houlsby (1995). The determination of the strain rate should be
based on the parameters required or the type of soil. Clarke (1995) pointed out that the
PMT modulus during the primary loading (e.g. the Ménard PMT modulus) does not
represent the elastic modulus in the ground because it is affected by the disturbance. It is
the unload-reload modulus that is more representative of the undisturbed modulus of the
soil.
Withers et al. (1989) has shown that in order to maximize the quality of the
unload-reload hysteresis cycle, a creep of the sand is necessary before the unloading
starts. This was confirmed by Clarke and Smith (1992), when they showed that the probe
continued to expand even if unloading was performed during a PMT on a mudstone at
depth 24 m. The creep will allow the sand to relax and the consolidation to take place.
Withers et al. (1989) recommended that for the Push-in PMT, the strain rate during the
creep in sands should be reduced to 0.1%/min before unloading. An illustration of the
procedure is fully explained by Withers et al. (1989) in their paper.
The duration for the creep should be around 10 min according to Clarke and
Smith (1995). Soilman and Fahey (1995) carried out most of the creep parts in their Self-
boring PMT on sands, in Perth, Australia, between 2 and 10 min. The creep time also
depends on the stress level.
104
The calibrations of the probe, the control unit, and the tubing in Push-in PMT are
usually performed before the actual PMT. The author suggests that the equipment
calibration should be carried out after the test. This is because the volume correction and
the pressure correction should be at a predefined stress level, which is the stress level
taken from the field test. The strain rate during the primary loading and the unload-
reload, and the final unloading should be kept the same when performing the calibrations.
Finally, the pressure and volume calibration trendlines are obtained by using the
Microsoft Excel application.
Vol
ume
Pressure
C1
C2
C4
C3
VOLUME CORRECTION PRESSURE CORRECTION
C4
Pres
sure
C1
Volume
C2
C3
CUR CUR
MathCadCubic root functions
ExcelCubic functions
Cell i = 1 to N
Pressure Pi, Volume Vi
Pi ≤ Pi+1
Trendline C3, C4 or CUR
Corrected Volume or Pressure
Trendline C1, C2 or CUR
YES: LoadingUnloading: NO
Figure 3-16. Proposed correction method in Push-in PMT.
105
While the Excel application provides a good trendline of cubic equation
Pressure = f(Volume) for the pressure correction, it may mislead as it provides a poor
trendline equation of the cubic root function Volume = f(Pressure), for the volume
correction (Figure 3-16).
Thus, for the Volume = f(Pressure) trendline equation, the author proposes the use
of the MathCad application where the use of a cubic root trendline equation is available.
Figure 3-16 also illustrates the shapes of the correction trendlines in Push-in PMT; a
flowchart used in the Excel spreadsheet to select the appropriate trendline equation at
different stages of the test is also presented.
Pressuremeter Soil Properties
It has been mentioned above that the PMT has its own basic parameters to
characterize the soil properties. They are: the strength parameter, limit pressure pl, and
the stiffness parameter, pressuremeter modulus usually denoted by EM. The
pressuremeter modulus EM is unique to the pressuremeter probe because it is derived
from the theory of elasticity (Hooke’s Law) in the cylindrical coordinate system.
Always, the assumption of a plane strain is applied (the vertical strain is zero), and
therefore the stress variations are related as follows
( )rz σσνσ θ ∆+∆×=∆ (3.60)
Making use of the equilibrium equation in this coordinate system, with the specified
boundary conditions, the shear modulus G of the soil is estimated as
VpVpG
v ∆∆
×=∆∆
=ε
(3.61)
Then conventionally, the mean pressuremeter modulus at the midpoint of the assumed
linear elastic range is
106
−
−
++=
of
ofofcM VV
ppVVVE
266.2 (3.62)
where the Poisson’s ratio ν = 0.33, Vc is the initial volume of the measuring cell, and pi
and Vi are as indicated in Figure 3-17. Expression of the pressuremeter modulus EM in
terms of the radial change can be found in Briaud (1992).
In foundation design, the PMT soil properties are widely used for bearing
capacity problems and settlement calculations. However, the cases of slope stability
analysis and sheet pile wall analysis, which require the Mohr-Coulomb based parameters
c and φ, cannot be achieved without using the methods such as the P-y curves (Baguelin
et al., 1978; Briaud, 1992).
Pres
sure
act
ing
on so
il
Volume change of cavity
A
B
Cpl
pf
po
voO vf
pl: Limit pressurepo: Soil initial pressure
pf: Creep pressure
Pseu
do-e
last
icPl
astic
Figure 3-17. Different parameters for defining PMT soil properties.
Correlations between the pressuremeter soil properties and c and φ, along with the
stiffness parameters E, have been developed. These correlations are the object of the
following section.
107
Correlating Soil Properties with PMT Data
Correlating the pressuremeter parameters would in turn classify the PMT as an
ordinary insitu test to determine the soil’s insitu stress state, strength, and stiffness. As it
will be seen, there are not many correlations encountered for the PMT. Most of the
existing correlations usually refer to the Ménard PMT soil properties. Fortunately,
comparative studies on the possible relationship between the parameters from the three
categories of PMT have been achieved. Baguelin et al. (1978) provided a comparative
table between the Ménard PMT and the Self-boring PMT data from Flanders Clay and
from Granitic Caudan Sand; Briaud (1992) also collected the data from Pre-bored PMT
(TEXAM PMT), Self-boring PMT, and Push-in PMT from the University of Houston,
and Texas A&M University, in sands and clays. It is thus possible to obtain at least a
rough approximation of the soil properties from any of the three types of PMT.
Soil classification
The table that is given to evaluate the rheological factor α in the settlement
calculation using the PMT method can be used for identifying the soil type with PMT.
Table 3-3. Soil type in PMT
Peat Clay Silt Sand Sand andgravel
Soil TypeEM/p*
l α EM/p*l α EM/p*
l α EM/p*l α EM/p*
l α
Over-consolidated >16 1 >14 2/3 >12 1/2 >10 1/3
NormallyConsolidated 1 9-16 2/3 8-14 1/2 7-12 1/3 6-10 ¼
Weatheredand/or
remoulded7-9 1/2 1/2 1/3 1/4
(Source: Baguelin et al., 1978)
108
The soil type presented in Table 3-3 depends on the ratio between the PMT
modulus EM and the limit pressure p*l, where p*
l = pl – po.Table 3-3 is about the best the
PMT can provide as far as soil classification is concerned.
Correlations for cohesionless and cohesive soils
Insitu state of stress. It has been discussed that the PMT creep pressure pf is
closely related to the preconsolidation stress cσ (Baguelin et al., 1978). Baguelin et al.
(1978) reported that Mori and Tama (1974) found out that the two parameters are equal
over a range of parameters. Similar findings were mentioned by Kulhawy and Mayne
(1990) for the case of Chicago lake clays. Several authors, including Kulhawy and
Mayne (1990) developed a correlation for clays, between the preconsolidation pressure
cσ and the limit pressure pl from Self-boring PMT, as follows
)98.0.. ;908.0 ;89( 45.0 2alc pDSrnp ====σ (3.63)
The coefficient of lateral stress, K0, was considered to be closely related to the
pressure po, but the hypothesis was unreliable and then abandoned (Baguelin et al.,
1978). The magnitude of the horizontal stress hσ depends on the type of probe, and it
was observed that only the Self-boring PMT can give good estimate of the horizontal
stress, hσ . Detailed explanation of difference between the initial pressure, po, and the
horizontal stress, hσ , is given by Clarke (1995). It is mentioned that the horizontal stress,
hσ , cannot be obtained from Push-in PMT curve because of the undefined extra-
horizontal stress acting on the membrane. However, a few methods have been developed
to identify hσ , for instance, lift-off method, method based on shear strength, curve fitting
method, etc. (Clarke, 1995).
109
Effective friction angle. Very few correlations have been established to estimate
the friction angle φ from the PMT basic parameters, pl, po, and/or EM. Briaud (1992)
mentioned that the prediction using the correlations can be very inaccurate. Clarke
(1995) reproduced the derived values of the friction angle φ from cone and pressuremeter
test, from various authors such as, Hughes et al. (1977), Robertson and Hughes (1986),
but they were qualified as not reliable. In 1970, the Centre d’Etudes Ménard and then
Muller, proposed a method that relates the friction angle φ with the PMT net limit
pressure (pl – po), but the correlation tends to significantly overestimate the value of the
friction angle. Researchers have used the curve fitting method to back-calculate the soil
properties of the soil. The curve fitting method is an iterative procedure by arbitrarily
selecting the input parameter in the simulation program, usually a finite element program,
in order to match the corrected PMT curve and the curve from the program by visual
observation. The curve fitting method has been used by researchers: Jefferies and Shuttle
(1995), Hicher and Michali (1995), and Altaee and Fellenius (1994). In this research, the
friction angle in the sand layers was determined by curve fitting using the PLAXIS code.
The finite element model with PLAXIS for the calculations is presented in Chapter 4 and
the resulting PMT fitted curves are given in Appendices B, D, and E.
Undrained cohesion. Theoretical calculations of the undrained shear strength su
of clays from PMT have been done (e.g., Bishop et al., 1945). Bishop et al. (1945) used
a method based on the net limit pressure, (pl – po). Other authors derived the same types
of equation as that from Bishop et al. Assuming the Poisson’s ratio of ν = 0.5, their
equations all reduce to
110
+=
−=
u
olu s
Gpps ln1 where β
β(3.64)
where G is the shear modulus of the clay. With the ratio G/su depending on the
overconsolidation ratio and typically ranging from about 100 to 600, the values of β in
Eq. (3.64) range from 5.6 to 7.4, Briaud (1992). Empirical relations based on PMT
database were also established by Briaud (1992) and other authors. A more refined
correlation is given below
( ) 75.067.0 olu pps −= (3.65)
Some other authors in Baguelin et al. (1978) (e.g., Amar and Jézéquiel, 1972) proposed a
linear relationship between the undrained cohesion su and the net limit pressure (pl – po).
Stiffness parameters. Baguelin et al. (1978) provided a discussion of the
difference between the PMT modulus and the laboratory triaxial compression test
modulus. They emphasized the argument by Ménard: not only that the stress paths
during the tests are different, but also the laboratory triaxial compression (drained) test is
a measure of the micro-deformation modulus, which is not the case for the Ménard PMT
modulus EM. Thus, the Centre d’Etudes Ménard (1975) proposed the use of Table 3-3 to
convert from the Ménard PMT modulus to the Young’s modulus in compression E; that
is Eq. (3.66) for Ménard PMT, and referring to Table 3-3:
αME
E = (3.66)
Following this method, which was selected for this research, the estimations of the
Young’s modulus from the Self-boring PMT and the Push-in PMT will make use of the
parameter relationships between the different methods of boring, discussed in the
111
beginning of this section. Using the data from Briaud (1992), the Young’s modulus from
Push-in PMT for clays from such conversion factor is given below.
α03.2ME
E = (3.67)
Advantages and Disadvantages of PMT
The main advantage of the PMT is its direct measurement of the initial horizontal
stress of the ground, especially the Self-boring PMT. As opposed to the SPT, the PMT is
less likely to be operator dependent. The main disadvantages of the PMT can be
attributed to the soil disturbance due to the method of insertion, limited access to gravels
and rocks and caving soils. While the Self-boring PMT is the most reliable PMT, it is
sometimes not practical because of the complication in the equipment operation, thus an
experienced operator is always required.
Summary on Insitu Testing
The four main insitu tests in the geotechnical engineering have been presented.
Each of them has its advantages and disadvantages in various conditions. The most
important subjects in evaluating the performance of the test should be based on the
accuracy and repeatability of the data, suitability to soil types (versatility), and the
equipment operation (simplicity and availability). Statistics on the characterization of the
performance of insitu tests from the analysis made by Orchant et al. (1988) can be found
in Kulhawy and Mayne (1990). In terms of reliability, they concluded that the electrical
CPT and the DMT are less variable than the PMT, whereas the SPT and the mechanical
CPT are the most variable tests. Furthermore, many comparison tables have been
produced in the literature on the applicability of each insitu test and accuracy on
estimation of the different soil properties (Lunne et al., 1997; Bowles, 1996; Clarke,
112
1995; Kulhawy and Mayne, 1990). The authors usually refer to the qualifications: A:
very good, or high; B: good, or moderate; C: not good, or low; D: poor, or not, etc.
Finally, Kulhawy and Mayne (1990) reproduced a comparative study on the cost of the
insitu tests that was developed by Handy in 1980, Figure3-18; his study was critical in
that he related the cost of the test with its accuracy.
Figure 3-18. Relative cost and accuracy of insitu tests (Source: Handy, 1980)
According the study by Handy (1980), among the insitu tests discussed in this chapter,
the Self-boring test is the most expensive and most accurate while the SPT and the
mechanical CPT are the least expensive and least accurate.
113
CHAPTER 4SHEET PILE WALL AT MOFFITT CANCER CENTER
Introduction
This chapter deals with one of the two full-scale tests that the University of
Florida performed as part of the research project financed by Florida Department of
Transportation (FDOT). Prediction of structural deflections (piles, footings, sheet pile
walls, etc.) in geotechnical problems is a better way to understand the relationship
between the real field performance and theoretical prediction. In this chapter, the lateral
deflections of a sheet pile wall are studied. The site is located at the campus of the
University of South Florida, Tampa, Figure 4-1.
Figure 4-1. Florida State and project entities.
114
Funding from the settlement of the State of Florida with the Tobacco companies
financed a new research tower as an addition to the Moffitt Cancer Center, which is part
of the College of Medicine, University of South Florida. Excavations were necessary for
the construction of the tower and an adjacent parking garage. Temporary shoring for
these excavations was to be provided by cantilevered sheet pile walls. The Moffitt
Foundation, overseeing the project, agreed to allow the University of Florida to perform
soil exploration tests at the site for the future wall.
Objectives
The insitu tests by the University of Florida were performed to be able to predict
the field performance: primarily the lateral deflections, of the sheet pile wall based upon
the data collected from: SPT, CPT and Pencel PMT (Push-in type of Pressuremeter).
Subsequently, comparisons and interpretations of the theoretical predictions with the field
actual deflections were undertaken.
Scope of Work
In order to accomplish the objectives mentioned above, the following tasks have
been included in this project:
• Collect the data from the SPT, CPT and PMT insitu tests and estimate the soilproperties for traditional method of analyses with CWALSHT, and numericalmodeling with PLAXIS.
• Instrument the sheet pile wall with Slope Inclinometer casings at 3 locations using2.5in.×2.5in. ×0.25in. inclinometer casings.
• Measure the actual deformation of the sheet pile wall before and after the excavation;
• Compare the inclinometer measured deflections with the theoretically predicteddeflections.
115
Site Description and Insitu Testing
Site Geological Characteristics
Around the area of University of South Florida, outcrop and shallow subcrop
rocks of Miocene age prevail. Middle and early Peistocene mineable limestone
constitutes the deeper layer around the Hillsborough County. Very often, the limestone
includes alluvial sinkholes. The cone penetrometer-derived soil stratigraphy by
Bloomberg et al. in 1988 at the University of South Florida campus indicates essentially
a sand layer near the ground surface, followed by silty layers, which are underlain by the
limestone (Randazzo and Jones, 1997).
Insitu Test Boring Locations
There was very little soil data for the area where the wall was to be installed. The
University of Florida conducted CPT and Pencel PMT at two locations adjacent to the
project wall. As indicated in the Field Exploration Plan of Figure 4-2, the wall is to be
constructed near the southwest corner in the area labeled INACCESSIBLE.
Figure 4-2. Locations of SPT, CPT, and Pencel PMT at Moffit Cancer Center.
116
The University of Florida cone truck was used to conduct the CPT and the Pencel
PMT at the locations indicated. LawGibb Group Member had already performed SPT
borings including the boring G-7 (on the north side and closest to the wall) selected for
this project.
Insitu Testing Data
The SPT boring log of the boring G-7 in Figure 4-2 was provided by LawGibb
Group Member and is presented in Figure 4-3. The shelby tube samples from the SPT
sounding permitted the identification of the soil type in the profile, which was essentially
made of fine medium dense sands (SPT N = 16) underlain by weathered limestones.
Figure 4-3 also shows the possible presence of karst mentioned in the geological
properties, from a depth of 30 ft to 40 ft below the ground surface (SPT N = 0). The
elevation of the ground surface at the boring G-7 was estimated to be around +40ft.
The CPT boring logs are presented in Figure 4-4 and the Figure 4-5 for the two sounding
locations CPT1 and CPT2, respectively. The estimated level of ground surface at the
CPT soundings was about +40 ft. The two CPT soundings display very good
repeatability of the data. The sand layer is marked by the higher values of the cone tip
resintance qc with a maximum recorded of qc = 250 bars.
At depth around 16.5 ft (5.0 m), the sleeve friction starts to increase which
indicates the clayey layers; below the clayey sand layers; at depth of about 32 ft (10.0 m),
the limestone is reached, the tip resistance dramatically increases and the test was stopped
at depth 34 ft (10.4 m) for CPT1 and 33 ft (10.0 m) for CPT2. Further soil properties
from the two CPT were obtained from the data reduction using the software Cpintr1 v.
3.04. The soil properties are presented in Appendix B in Tables B-1 and B-2.
117
Figure 4-3. SPT boring log and soil stratigraphy at Moffitt Cancer Center.
The PMT soundings also show consistency of the results between the soundings
PMT1 and PMT2, Figure 4-6 through Figure 4-9. The steep slope of the primary loading
in the first 15 ft (4.5 m) is indicative of the medium dense sand layer; below the sand
layer, the PMT curves, for instance, at depth 25 ft (7.6 m) become flat, which is
characteristic to clay type layers. The maximum depth reached with the PMT was 25 ft
(7.6 m), which was sufficient regarding the geometrical properties of the sheet pile wall
that will be discussed in the next subsection. Some of the slopes of unload-reload part of
the PMT curves are apparently negative, due probably to the lack of creep before the
unloading process
118
Figure 4-4. CPT sounding CPT1 (South) at Moffitt Cancer Center.
119
Figure 4-5. CPT sounding CPT2 (North) at Moffitt Cancer Center.
120
Moffit PMT 1
0
2
4
6
8
10
12
14
16
-10 0 10 20 30 40 50 60 70 80 90 100Volume (cm 3)
Pres
sure
(bar
s)5ft
10ft
15ft
Figure 4-6. Corrected PMT curves: PMT1 (South) at Moffitt Cancer Center.
Moffit PMT 1
0
2
4
6
8
10
12
14
16
-10 0 10 20 30 40 50 60 70 80 90 100Volume (cm 3)
Pres
sure
(bar
s) 20ft
25ft
Figure 4-7. Corrected PMT curves: PMT1 (South) at Moffitt Center.
121
Moffitt PMT 2
0
2
4
6
8
10
12
14
16
-10 0 10 20 30 40 50 60 70 80 90 100
Volume (cm 3)
Pres
sure
(bar
s) 5ft10ft15ft
Figure 4-8. Corrected PMT curves: PMT2 (North) at Moffitt Center.
Moffitt PMT 2
0
2
4
6
8
10
12
14
16
-10 0 10 20 30 40 50 60 70 80 90 100
Volume (cm 3)
Pres
sure
(bar
s)
20ft
25ft
Figure 4-9. Corrected PMT curves: PMT2 (North) at Moffitt Center.
122
Sheet-Pile Wall Test Section
As mentioned previously, a cantilevered sheet pile was installed to shore up the
excavation of the parking garage. The sheet pile section selected by the Contractor,
Turner, Inc., was the CZ128. Regarding the depth of excavation and a quick design
check on the sheet pile wall for the site, the CZ128 appeared too rigid for the project.
The cantilevered sheet piles were driven until reaching the weathered limestone. They
were to be embedded approximately a foot into the weathered limestone to secure the
stability. This procedure results in pile tip elevations to vary between +10.0 ft and
+25.5 ft (according to the original driving log datasheet provided by the Ardaman and
Associates, Inc.). A typical drawing of the profile with the driven sheet piles is shown in
Figure 4-10.
Figure 4-10. Soil-cantilevered sheet pile wall profiles at Moffitt Cancer Center.
123
In order to monitor the wall lateral deflections along the depth of the wall, a slope
inclinometer was used. Instead of the traditional circular grooved casings, steel box
sections were welded to the back of the sheet piles and driven with a piece of CZ128 at
three locations along the wall. Figure 4-11 shows a cross section of the casing attached
to the CZ128 pile. The inclinometer probe was inserted along the diagonals of the box
section and reads the deflections at 45º offset to the perpendicular to the wall. The
software for inclinometer data reduction, the DMMWin Version 1.1.0, can correct the
readings for the offset and provide the deflections in two directions: perpendicular to the
plane of the wall and parallel to the wall (A+ and B+ directions, respectively).
Figure 4-11. Sheet pile section CZ128 with 2.5 in.×2.5 in.×0.25 in. casing.
The photos in Figure 4-12 were taken at the Moffitt site showing the pile driving
process, Figure 4-12A, and one of the three inclinometer casings location along the wall,
Figure 4-12B. The ground surface elevation behind the wall within the three
inclinometer casings is rather uniform. The plan view of the concerned wall is shown in
Figure 4-13 where the ground surface elevations at the casings are indicated: +42.77 ft,
+40.94 ft and +39.52 ft. After the excavation, the bottom of the excavation is level and is
124
at Elevation +31.5 ft for the three casings, and the elevation where the parking area is
designed, at Elevation +25.5 ft.
A B
Figure 4-12. Driving of section CZ128 and location of inclinometer casings. A) Drivingof CZ128. B) Casing welded to CZ128.
The casing 2 in Figure 4-13 is the casing that would most probably meet the plane
strain criteria for the modeling in the traditional method and in the finite lement method
with PLAXIS. It was confirmed with the fact that the top deflection of the wall at casing 2
was the largest in magnitude among the three casings used. Thus, the predictions, for
both traditional method and finite element method were focused on the deflection of the
wall at casing 2. The bottom of the wall at that section is at Elevation +25.5 ft, and the
final soil-structure profile for the theoretical analyses is presented in Figure 4-14.
125
N
Figure 4-13. Plan views of sheep pile wall at Moffitt Cancer Center site.
Soil-Structure Profile for CWALSHT and FEM Modeling
Modeling of the soil profile along with the wall was done at the section of the
wall corresponding to casing 2. As a reminder, the bottom of the wall is slightly
embedded in the weathered limestone at the bottom of the profile. Consequently, for the
numerical solution in PLAXIS, a pin support is installed at the bottom of the wall in order
to prevent any lateral deflection of the bottom. The assumption in the conventional
method already matches the particular case of Moffitt Cancer Center sheet pile wall; that
is the wall is fixed at its bottom. Figure. 4-14 shows the final profile along with the soil
stratigraphy from the CPT1 sounding, the exception is then made by the existence of the
126
limestone at the bottom of the wall. The total length of the wall at section of casing 2 is
14.5 ft (L = 39.5 – 25.0 = 14.5 ft).
Figure 4-14. Final soil-structure profile for theoretical predictions.
Slope Inclinometer Data
The Digitilt Inclinometer and its accessories were provided by the Florida
Department of Transportation (FDOT). The device consists of three major components:
the casing sensor which is equipped with two pairs of guide wheels, Model 50325E; the
digital indicator, Model 50309, Serial Number, S/N 26084; and the interconnecting
electrical cable (with pulleys). The square inclinometer casings used are made of steel
and were much stiffer than the conventional grooved circular plastic or aluminium
casings. Each one was welded to a sheet pile before the driving with a sharp shoe at its
127
toe. Because of the difference from the conventional inclinometer conditions, a brief
description of the field measurements is needed.
Procedure During Measurements
The type of inclinometer equipment system reads the data every 1.0 ft of depth
along the casings. The data were taken in the two opposite directions with the guide
wheels sliding along the diagonals of the steel casings. Thus, the A0 and B0 readings
were taken when the guide wheels made 45° with the direction parallel (or perpendicular)
to the plan of the CZ128 wall, that is pointing to the northeast. The A180 and B180
readings were taken when the sensor is oriented 180° from the probe direction when A0
and B0 readings were taken (probe pointing to southwest). The upper guide wheel of
each pair defines the direction of the sensor as it is indicated in Figure 4-15 below.
Adding algebraically the values of A0 and A180, on one hand and the values of B0 and
B180, on the other hand, ensured the checking of the validity of the data collected in the
field. The results from the algebraic addition were found to be consistent values around
15, which indicates the validity of the readings.
Two sets of readings were taken: the first ones were before the excavation, it is
the initial reading in the inclinometer field test; the second readings were taken after the
excavation is finished, for example, the profiles shown in Figures 4-10 and 4-14. The
effect of creep in the soil was not considered in this research. The readings A and B are
the basic data that were further reduced in order to obtain the cumulative lateral
deflection of the wall.
128
Guide wheels
Electrical cable
Steel casing
Diagonal of 2.5in×2.5in×0.25in
steel box
A B
Readings on the Digital indicator
Switch button
CZ128 wall plane
A0 reading
B0
read
ing
A180 reading
B18
0 re
adin
g
Dire
ctio
n of
pro
cess
1 foot marks for reading
2.0f
t
Figure 4-15. Procedure taking readings A and B with steel box casing.
Data Reduction with DMMWin Program
Among the existing softwares for data reduction, the Digitilt Datamate Manager for
Windows was selected mainly because of the availabitlity of the freeware: DMMWin,
Version 1.1.0, manufactured by Slope Indicator Company. The data reduction is divided
into two parts. The first part consists of input with the calculations presented in tabular
form with the specification that the inclinometer sensor was oriented at 45° from the
perpendicular to the wall. The final cumulative deflections at 1-foot interval of depth
was obtained as final output, Tables 4-1 and 4-2. The second part is the plots of the
lateral deflections in profile view (deflection versus depth), and in plan view, as shown in
Figures 4-16 and 4-17, respectively. It should be noted that the data reduction with the
129
DMMWin software allows obtaining only the deflection up to 2 ft below the opening of
the casing, which is the last reading from each direction during the field measurement
(the guide wheels having to be kept inside the casing). The top deflections of the wall
were then calculated by linear extrapolation of the last two deflections in the Tables 4-1
and 4-2, (depths 2 ft and 3 ft). The lateral maximal deflection recorded was rather small:
0.14 ft at depth 2 ft below the ground surface, which resulted in the lateral top deflection
of 0.17 ft (at ground surface elevation) after linear extrapolation. A backup computation
on the Excel spreadsheet was also prepared and summarized in Table 4-3 for verification.
The results from the spreadsheet computations obtained agree well with those from the
DMMWin program. The procedure followed to obtain the results in Table 4-3 is given in
Appendix B. As a result of the imperfection in the field (e.g., perfect plane strain
scheme, symmetric position of the soil-structure system, etc.), deflections in the direction
parralel to the sheet pile wall developed. Fortunately, the magnitude of such deflection
was negligible as the maximum recorded was 0.06 in. at depth 2 ft below the ground
surface.
Table 4-1. Input and output from DMMWin - A readings on casing 2
130
Table 4-2. Input and output from DMMWin - B readings on casing 2.
Figure 4-16. Lateral deflection and parellel direction versus depth (casing 2).
131
Figure 4-17. Total deflection in plan view (casing 2).
Table 4-3. Excel spreadsheet for inclinometer testing (casing 2)
A Axis Data Reduction
Depth (ft) Init. Diff. A+ A- Alg. Diff. CHANGE Deflect. Cumul. (in.)
13 -261.5 -104.5 122 -226.5 35.0 0.011 0.01112 -203.5 -73 92.5 -165.5 38.0 0.011 0.02211 -120.0 -34.5 52.5 -87 33.0 0.010 0.03210 -44.5 -0.5 14.5 -15 29.5 0.009 0.0419 3.0 30.5 -11 41.5 38.5 0.012 0.0528 36.5 50.5 -32 82.5 46.0 0.014 0.0667 76.5 71.5 -51 122.5 46.0 0.014 0.0806 100.0 80.5 -60.5 141 41.0 0.012 0.0925 103.0 81 -65.5 146.5 43.5 0.013 0.1054 103.0 83.5 -64 147.5 44.5 0.013 0.1193 95.0 76.5 -56 132.5 37.5 0.011 0.1302 84.0 73 -52 125 41.0 0.012 0.142
Modeling with CWALSHT and FEM Code PLAXIS
As it has been discussed in Chapter 2, different sets of input parameters are
required for the two methods. Within the finite element code PLAXIS alone, different
constitutive models require different sets of soil parameters for input. However, for the
132
soil properties, the input parameters for both methods and the constitutive models in
PLAXIS were obtained from the same correlations. These correlations were introduced in
Chapter 3 and are repeated here for convenience. The soil stratigraphy for CWALSHT
and PLAXIS are the same from each type of insitu test but the modelings are
fundamentally different.
Correlations for Soil Properties
The unit weights were estimated from the FLPier Manual (FDOT, 2001).
Considered as not having significant impact on the prediction results, the unit weight
correlations were used for all other insitu test input of this research.
In addition to the actual SPT boring G-7, equivalent SPT N blow counts from the
CPT were also included in the computations; these are indicated as CPT (N) throughout
the rest of the work. For the SPT and the CPT based blow counts CPT (N), the friction
angles were correlated from Peck et al. (1974).
( )N×−×−=° 0147.0exp6034.27881.53][φ (4.1)
The undrained shear strength su was obtained from Terzaghi and Peck, (1948). The
Young’s Modulus was correlated from Eq. (4.2) that was provided by Bowles (1996):
( )( ) Claysfor 200 to150tsf][
Sandsfor 15500[kPa]
usENE
×=+×= (4.2)
For the CPT, because the correlation from Robertson and Campanella (1983)
greatly overestimates the effective axisymmetric friction angles for the finite element
modeling with PLAXIS, the more conservative correlation with values closer to triaxial
friction angle from Kulhawy and Mayne (1990) was selected.
133
+=°5.0
0
log0.116.17][
a
v
a
c
p
pq
σφ (4.3)
where 0vσ is the effective overburden pressure and pa, the atmospheric pressure.
The undrained shear strength su from CPT is based on the following empirical
correlation
15 to10 where 0 =−
= kk
vcu N
Nq
sσ (4.4)
As for the Young’s Modulus E, the following Eq. (4.5) from Bowles (1996) were
used to correlate with qc:
( )( ) Claysfor 8 to3
sandsSilty and Sandsfor 4 to2
c
c
qEqE
×=×= (4.5)
The friction angle values from the PMT were estimated from the curve-fitting
method. The curve-fitting method is achieved with the finite element code PLAXIS. A
special finite element model is aimed to match the corrected PMT curves from each depth
where a PMT was carried out in the field. Figure 4-18 shows the triangular 6-noded
element meshing, the overburden load A-A and distributed load B-B from the probe cell,
and the boundary conditions. A typical result from such curve fitting method, using both
Mohr-Coulomb model and Hardening Soil model is presented in Figure 4-19.
The calculation process for the typical result of Figure 4-19 consists of three phases: the
first phase is a Plastic type of calculation where the overburden load A-A is applied; after
the displacement at the end of first phase is reset to zero, the second phase is started by
applying the pressuremeter cell pressure B-B with the Updated mesh type of calculation;
the third hase is the unloading phase where the magnitude of the pressure B-B is reduced,
134
this phase is also under the Updated mesh type of calculation; the third hase is the
unloading phase where the magnitude of the pressure B-B is reduced, this phase is also
under the Updated mesh type of calculation. In order to convert from the radial
displacement obtained from PLAXIS to the volume change in the actual PMT, the equation
below is used.
00 r
hVV
r −×
∆+=∆
π(4.6)
where, V0 is the initial volume of the measuring cell; ∆V, the change in volume of the
measuring cell; h, the height of the measuring cell; and r0, the initial radius of the
assumed cylindircal shape of the cell. The curve-fitting results for the other PMT are
provided in Appendix B.
Figure 4-18. Curve-fitting model (Unit: m).
135
Figure 4-19. PMT curve-fitting with PLAXIS.
The undrained cohesion is directly computed from the correlation provided by
Baguelin et al., (1978), or Briaud (1992).
βol
upp
s−
= (4.7)
where β = 6.5 was chosen from the boundary values: 5.6 and 7.4.
The stiffness parameters from the PMT were estimated using the relationships
between different PMT in Briaud (1992). In order to convert the parameters from the
Push-in Pencel PMT modulus to the standard triaxial compression test modulus, the
Pencel PMT modulus is divided to the conversion factors 2.03 and α as discussed in
Chapter 3 (Baguelin et al., 1978). The factor 2.03 was particularly for clays, however
due to small number of data when obtaining the conversion factor, it is considered to be
more conservative and thus safer to use for all soil type.
α×=
03.2IN-PUSHEE (4.8)
where α is a function of the ratio EM/pl, Table 3-3, Chapter 3.
136
Modeling with CWALSHT
The linear elastic behavior of the CZ128 is assumed in CWALSHT. The
parameters required are the steel Young’s modulus E and the moment of inertia I. The
properties are presented in Table 4-4 and compared with those required for PLAXIS.
The main soil properties are the strength parameters and the stiffness parameters.
As supplementary data, the soil properties from the Miniature Cone Penetration Test
(MCPT), performed by the Ardaman and Associates were included in the analysis with
CWALSHT. The equivalent SPT N blow counts from sounding C-3 of the MCPT were
used (Appendix B). A typical profile of the soil stratigraphy with the structure is
presented in Figure 4-20. The steep slope on the excavation side of the wall was
achievable. The wall friction and wall adhesion were calculated in the following manner
Sandsfor tan21tan wall φφ = (4.9)
Claysfor 21
wall usc = (4.10)
The soil properties from insitu tests are presented in Table 4-5 through Table 4-7.
Figure 4-20. Soil-structure profile type for CWALSHT analysis (Unit: ft).
137
Table 4-4. CZ128 wall properties for CWASHT and PLAXIS
Table 4-5. Soil properties from MCPT for CWALSHT (Ardaman and Associates)
Layer MCPT N Unit Weight φ suWall
FrictionWall
AdhesionBottom (ft) (bl/ft) (pcf) (o) (psf) (o) (psf)
GS*=0
12.008 112 33 - 14 -
15.00
20 118 36 - 14 -
20.00
15 118 36 - 14 -
26.00
25 118 36 - 14 -
30.00
10 112 33 - 14 -
36.0010 110 - 1100 - 600
EOB**
- 135 - 2000 - 1000
*GS =Ground Surface: Elev. +42.5ft**EOB =End of Boring
Table 4-6. Soil properties from SPT boring G-7 for CWALSHT analysis
Bottom SPT N γ φ WallFriction
(ft) (bl/ft) (pcf) (o) (o)GS*= 0
4.75 4 99.9 27.9 14.8
7.25 10 115.1 30.1 16.1
11.00 16 117.7 32.1 17.4
16.00 14 116.4 31.4 17.0
21.00 7 102.4 29.0 15.5
26.00 5 101.1 28.2 15.0*S =Elevation +39.5ft
CWALSHT PLAXIS
Cross Section Area, A – EA = 2.227E+08 (lb/ft)Elastic Modulus, E 29,000,000 (psi)
Moment of Inertia, I 236.5 (in.4) EI = 4.763E+07 (lb.ft2/ft)
Equivalent Thickness, d – 1.602 (ft)
Weight, w – 26.2 (ft/ft)
138
Table 4-7. Soil Properties from soundings CPT1, PMT1 for CWALSHT
CPT (N) CPT PMT
Bottom SPT N γm φ suWall
Friction Adhesion qc φ suWall
Friction Adhesion EM pl φ suWall
Friction Adhesion
(ft) (bl/ft) (pcf) (o) (psf) (o) (psf) (psi) (o) (psf) (o) (psf) (psi) (psi) (o) (psf) (o) (psf)
GS*=0 36 114.5 37.6 - 22.5 - 2580.1 47.8 - 28.2 - 4.92 2157.3 214.6 37.8 - 22.7 -7.51 14 111.3 31.4 - 18.8 - 857.2 38.8 - 25.8 -
12.30 1880.5 166.8 33.5 - 20.1 -12.50 17.49 1449.9 40.0 - 27.0 - 1982.4 152.3 31 - 18.6 -18.86 25 114.5 34.8 - 20.9 - 19.69 1632.8 181.3 30 - 18.0 -22.51 17 120.9 - 2610 1305 22.97 326.3 - 2944 - 1472 33.63 8 114.5 - 1044 522 2214.5 116.0 - 1650 - 824.8
EOB**
*GS = Ground Surface: Elevation +40.0ft
**EOB = End of Boring
139
Modeling with PLAXIS
The sheet pile wall problems in the finite element modeling are a plane strain
problems. The 15-noded triangular finite elements were selected over the soil-structure
profile shown in Figure 4-21. The choice of the 15-noded elements allowed the coarser
meshing, however the meshing around the wall is refined to a factor of 2. Interface
elements are installed around the wall in contact with the soil. The reduction factor of the
interface was Rf = 0.7 for friction between sands and the CZ128 wall, and Rf = 0.5
between clays and the CZ128 wall. The bottom of the wall is pinned as to simulate the
embedment of the wall in the limestone. During the calculation process, the excavation
by steps is performed in three stages, each stage to remove about 2 to 3 ft depth of soil.
Table 4-4 above presents the properties of the wall for input parameters in
PLAXIS. The soil input parameters from the SPT boring G-7 are given in Table 4-8
whereas those from the CPT and PMT are in Tables 4-9 and 4-10 for the Mohr-Coulomb
model and the Hardening-Soil model, respectively.
The soil input parameters for the curve-fitting method resulting from Fig 4-19 are
also presented in Table 4-11. The stress dependent stiffness of the soil can be observed
with the Hardening-Soil model. While the friction angles are in the same order of
magnitudes from the two constitutive models, the moduli are quite different, especially in
the first layer. Compared to the modulus from the insitu test, for instance, Table 4-10, the
moduli from the curve-fitting method are unrealistically high. Such values were not used
in the prediction calculations with in PLAXIS.
140
Figure 4-21. Finite element modeling with PLAXIS (Unit: ft).
Table 4-8. Soil properties from SPT boring G-7 for PLAXIS
Bottom SPT N γ φ Es Eur=3×Es
(ft) (bl/ft) (pcf) (o) (psi) (psi)GS = 0
4.75 4 99.9 27.9 1377.5 4132.5
7.25 10 115.1 30.1 1812.5 5437.5
11.00 16 117.7 32.1 2247.5 6742.5
16.00 14 116.4 31.4 2102.5 6307.5
21.00 7 102.4 29.0 1595.0 4785.0
26.00 5 101.1 28.2 1450.0 4350.0GS =Elevation +39.5
CZ128 Wall
Excavation side, 3 stages
1
2
3
Pin support at the bottom
Interface Elements
141
Table 4-9. Soil properties from CPT1 and PMT1 for Mohr-Coulomb model PLAXIS
Mohr-Coulomb ModelCPT (N) CPT PMT
Bottom γm φ su Eref φ su Eref φ Eref c Bottom(ft) (pcf) (o) (psf) (kPa) (o) (psf) (psi) (o) (psi) (psf) (m)
GS*=0 114.5 37.6 - 3697.5 47.8 - 5118.5 GS=04.92 37.8 3188.1 - 1.50
7.51 111.3 31.4 - 2102.5 38.8 - 2504.3 2.29
12.30 33.5 2779.1 - 3.75
12.50 3.81
17.49 40 - 4893.3 31 2929.7 - 5.33
18.86 114.5 34.8 - 2900.0 5.75
19.69 30 2437.5 - 6.00
22.51 120.9 - 2610 2718.8 6.86
22.96 - 2944 1794.4 7.00
33.63 114.5 - 1044 1450.0 20 1636.0 1650 10.25
EOB** EOB *GS = Ground Surface = Elev. +40ft, c’ and φ’ from curve-fitting with M-C model**EOB = End of Boring
Table 4-10. Soil properties from CPT1 and PMT1 for Hardening-Soil model PLAXIS
Correlated for Hardening Soil:m =0.5, νur = 0.2, Eoed = E50
ref, Eur = 3×E50ref
CPT(N) CPT PMTBottom Ε50
ref φ su Ε50ref φ su E50
ref φ Bottom(ft) (psi) (o) (psf) (psi) (o) (psf) (psi) (o) (m)
GS*=0 3697.5 37.6 - 5118.5 47.8 - GS=04.92 3188.1 38.0 1.50
7.51 2102.5 31.4 - 2504.3 38.8 - 2.29
12.30 2779.1 31.0 3.75
12.50 3.81
17.49 4893.3 40 - 2929.7 30.0 5.33
18.86 2900.0 34.8 - 5.75
19.69 2437.5 29.0 6.00
22.51 2718.8 - 2610 6.86
22.97 1794.4 - 2944.1 7.00
33.63 1450.0 - 1044 1636.0 19.0 10.25
EOB** EOB*GS = Ground Surface = Elev. +40.0ft
**EOB =End of Boring
142
Table 4-11. Soil properties from PMT1 curve-fitting with PLAXIS
Mohr-Coulomb Hardening SoilLayerBottom
(ft)Eref
(psi)φ(o)
c(psi)
E50ref
(psi)φ(o)
GS=0 4.92 21895 37.8 - 42050 38.07.51
12.30 21025 33.5 - 29000 31.012.50 17.49 20155 31.0 14500 30.018.86 19.69 18850 30 - 10295 29.022.51 22.97 33.63 10585 20.0 104.4 4350 19.0EOB
Results from CWALSHT
The top deflection of the CZ128 wall from each insitu test data were compared to
each other. Table 4-12 shows that all the insitu tests underpredict the top deflection of
the wall by at least a factor of two. The range of values is small; this is probably due to
the assumption that the bottom of the wall is fixed and no rotation is allowed.
Table 4-12. Wall top deflection from CWALSHT analysis
Insitu Testing Top Deflections(in.)
Measured 0.17
MCPT 0.07
SPT 0.07
CPT (N) 0.06
CPT 0.04
PMT 0.06
143
Results from PLAXIS
The top deflections from the two constitutive models: the Mohr-Coulomb model
and the Hardening-Soil model are given in Table 4-13. The results show larger range of
values of the magnitudes than in CWALSHT, with the largest top deflection of 0.72 in.
from the Mohr-Coulomb SPT and 0.03 in. from the Hardening Soil CPT.
Table 4-13. Wall top deflection from finite element method with PLAXIS
Top Deflections(in.)
Measured 0.17
Predicted Mohr-Coulomb Hardening Soil
SPT 0.72 0.18
CPT(N) 0.54 0.14
CPT 0.11 0.03
PMT (Correlated) 0.56 0.13
PMT (Curve fitting) 0.10 0.05
Discussion on Finite Element Modeling
Influence of the Choice of Modulus and Constitutive Model
The different strength parameters from the various insitu tests enabled an analysis
on the effect of these strength parameters on the results. In particular, the friction angle,
φ, with the finite element modeling; the use of the elastic modulus E in the finite element
program was also studied. In the Mohr-Coulomb computations, the moduli used are the
loading moduli from each insitu tests; the Hardening Soil model includes the unload-
reload modulus. The soil around the sheet pile wall undergoes unloading conditions after
the excavation: the backfill side in active state, and the bottom of excavation tending to
144
heave. In this study, computation using the unload-reload modulus was also carried out.
The unload-reload moduli Eur are estimated to be three times the reference moduli as
suggested in the PLAXIS manual (three to five times the reference moduli) and listed in
Table 4-10:
Eur = 3×Eref
The top deflections of the Mohr-Coulomb model with only the reference modulus
E as input are given in Table 4-14. Inputting simply the reference modulus E, Mohr-
Coulomb model does not take into account the unloading effect of the soil, it was realized
that using the unload-reload modulus Eur was more appropriate. Consequently, a
reanalysis using the unload-reload modulus in the Mohr-Coulomb model was performed.
The resulting deflections at the top of the wall are listed in Table 4-15 for comparison
with the measured deflections and the previous calculations. The comparison shows that
the Mohr-Coulomb model using the unload-reload modulus is very close to the
Hardening Soil model in the cantilevered sheet pile wall analysis.
Linear Elastic Model
When constructing a structure such as a retaining wall, the designer would always
want the wall to be safe with a reasonable factor of safety, say from 1.5 to 3.0. Thus, the
soil does not undergo large deformations and hence the stiffness involved is constant and
is equal to the Young’s modulus. A computation with the finite element Linear Elastic
model of the soil was performed. This model uses as main input parameters only the
moduli in Table 4-9. The strength parameters c and φ are not involved in the
computation. As shown in Table 4-14, the resulting predicted top deflections are
significantly smaller than those from Mohr-Coulomb model and Hardening Soil model.
145
Measured versus Predicted Deflections
Figure 4-22 through Figure 4-24 present the plots of the inclinometer measured
deflections versus depth with all of the theoretically predicted deflections.
Table 4-14. Top deflections of the wall: measured versus predicted
1. Measured with Slope Inclinometer
0.17Deflection (in.)
2. Predicted with Traditional method and FEM
PMT*(φ curve-fitted)Calculation Methods MCPT SPT CPT(N) CPT
Correlations Curve Fitting
CWALSHT 0.07 0.07 0.06 0.04 0.06
PLAXISMohr-Coulomb - 0.72 0.54 0.11 0.56 0.09
PLAXISHardening Soil - 0.18 0.14 0.03 0.13 0.05
PLAXISLinear Elastic - 0.10 0.08 0.03 0.07
*curve-fitting, φ constant, E varied
Table 4-15. Effect of modulus and constitutive models with PLAXIS
1. Measured with Slope Inclinometer
0.17Deflection (in.)
2. Predicted with FEM using PLAXIS
Test Model & Moduli SPT CPT(N) CPT PMT
Mohr-CoulombModulus Eref 0.72 0.54 0.11 0.56
Mohr-CoulombModulus Eur = 3×Eref 0.27 0.15 0.007 0.16
Hardening SoilModulus E50
ref 0.18 0.14 0.03 0.13
146
25
27
29
31
33
35
37
39
0 0.1 0.2 0.3 0.4 0.5 0.6
Wall Deflection (in)
Ele
vatio
n (f
t)Measured
CWALSHT (SPT)
CWALSHT (CPT)
Plaxis MC (SPT)
Plaxis MC 3xE (SPT)
Plaxis HS (SPT)
Plaxis MC (CPT)
Plaxis MC 3xE (CPT)
Plaxis HS (CPT)
Figure 4-22. Deflection versus depth using SPT N and CPT (N) blow counts.
25
27
29
31
33
35
37
39
0 0.1 0.2 0.3 0.4 0.5 0.6Wall Deflection (in)
Ele
vatio
n (f
t)
Measured
CWALSHT
CWALSHT (MiniCPT)
Plaxis MC
Plaxis MC 3xE
Plaxis HS
Figure 4-23. Deflection versus depth using CPT data.
147
25
27
29
31
33
35
37
39
0 0.1 0.2 0.3 0.4 0.5 0.6
Wall Deflection (in)
Ele
vatio
n (f
t)
Measured
CWALSHT
Plaxis MC
Plaxis MC 3xE
Plaxis HS
Figure 4-24. Deflection versus depth using PMT data.
All of the wall top deflection predictions by CWALSHT (average: 0.060 in.) are
less than half the actual measured deflection of 0.167 in.. Also, CWALSHT is not very
sensitive to the variation of the strength parameters c and φ; for instance, despite the large
range of the friction angle values between CPT (38.8° to 47.8°) and SPT (27.9° to 32.1°),
the difference in deflections from using CPT and SPT derived input parameters is only
about 0.02 in.. Furthermore, CWALSHT does not require any stiffness parameters of the
soil as input. A review of the insitu test input shows that the friction angles from SPT,
CPT (N), PMT and MCTP are all in the neighborhood of 32.0°, those from CPT are
around 45.0°. The highest input value accepted by CWALSHT was 47.0°. Below
Elevation +34.0 ft (depth around 6.0 ft), the undrained cohesion also varies significantly
among the four tests: ranging from 2949 psf (CPT) to 1100 psf (MCPT). Another
148
computation was performed by keeping all other parameters from the CPT the same, and
just varying the cohesion values; the results showed that the effect is negligible.
The strength parameters c and φ used for the finite element analysis are the same
as those for CWALSHT except for the Hardening Soil modeling using PMT data. It was
observed that the modulus of the soil Eref or E50ref has a fairly large influence on the
results for both Mohr-Coulomb model and Hardening Soil model. In addition, the
change in friction angle value also affects the prediction of the wall deflections. For the
case of Mohr-Coulomb model, considering the same order of magnitude of the stiffness
from SPT, CPT (N), CPT and PMT; the predicted top deflection using the high friction
angle values from CPT data, 47.8°, was 0.11 in. which is about 80% less than that
resulting from using the average friction angles of 33.0° from the PMT data (0.56 in.). It
is also encountered for the case of the Hardening Soil model: 0.03 in. versus 0.13 in..
For the PMT parameters, if the stiffness values obtained from the curve-fitting
method, which are considerably higher (Table 4-11) were used, unrealistic results were
obtained. In the case of the Mohr-Coulomb model, the top deflection was under-
estimated by about 50%: 0.09 in.. Thus, a change in modulus from the correlated ones
(Table 4-9) to curve-fitted ones (Table 4-11) in the Mohr-Coulomb model lowered the
prediction from 0.56 in. to 0.09 in.. For the case of Hardening Soil model, the curve-
fitting moduli values resulted in a top deflection about half of that from Mohr-Coulomb
model: 0.05 in. and 0.09 in., respectively.
Computations with the Mohr-Coulomb model using the unload-reload moduli Eur
in Table 4-10 give good predictions using the input derived from SPT, CPT(N) and PMT.
Good predictions were obtained as well by using the reference moduli E50ref with the
149
Hardening Soil using parameters from SPT, CPT (N), and PMT. These results are shown
in Table 4-15.
In summary, the plots in Figure 4-22 through Figure 4-24 can be divided in 3
groups of finite element method predictions: (1) the smallest deflections (both CPT), (2)
the more accurate predictions, and (3) the overpredicted deflections. The underpredicted
deflections are due to the high friction angles. The unloading conditions in the problem
are considered by Mohr-Coulomb only by introducing directly the unload-reload modulus
Eur, whereas it is taken into account by Hardening Soil model through the reference
modulus E50ref.
Conclusions
The results from this analysis enabled to draw the first conclusions on the finite
element method with PLAXIS:
Higher friction angle produces less deformation;
Higher soil stiffness produces less deformation;
Mohr-Coulomb model results in overly higher values of the deflection than those
from the Hardening Soil model, because the latter uses the unload-reload modulus;
Using the unload-reload modulus Eur in the Mohr Coulomb model can be as good
analysis as Hardening Soil model in the present cantilevered sheet pile wall.
Concerning the input parameters for the finite element analysis: the most accurate
predictions are those obtained from the SPT and CPT (N), and PMT (correlated) derived
parameters using the Hardening Soil model and the Mohr-Coulomb model with unload
reload modulus, followed by the CPT parameters using the Mohr-Coulomb model. The
common point of these three scenarios is that they all have the same order of magnitude
of the elastic modulus; the high friction angle from CPT overcompensated the softer
150
Mohr-Coulomb loading modulus. For the other predictions using the simple Mohr-
Coulomb model, it iwas noted that the strength and the stiffness parameters, and the
models did not compensate to obtain better results.
The results obtained from the Linear Elastic model for soils showed that generally
the predicted deflections are smaller than in Mohr-Coulomb and in Soil Hardening
models, Table 4-14. Thus, the Linear Elastic model is too unconservative even though
the measured deflection was fairly small.
The following conclusions can be stated:
The friction angle values from CPT using the correlation by Kulhawy and Mayne
(1990), Eq. (4.3), are still high for the traditional method and the finite element method
although they are generally smaller that those from the correlation Robertson and
Campanella (1983);
The soil properties from SPT and CPT (N), and those correlated PMT, Baguelin et
al. (1978) and Briaud (1992) give better results for the Hardening Soil model with the
friction angles evaluated from correlation of Eq. (4.1), and the curve-fitted friction
angles, respectively;
It was found that even at very small deformation of the soil, the Linear Elastic
model for soils is not appropriate for analysing sheet pile wall problems;
The traditional method using CWALSHT underpredicts wall deformations
unconservaively and has less sensitivity on the input parameters;
The finite element code PLAXIS has more analysis capability and parameter
sensitivity over CWALSHT;
151
For unloading problems, the unload reload modulus Eur can be reliably used with
the Mohr-Coulomb model.
152
CHAPTER 5STUDY OF OTHER SHEET PILE WALLS
General
This chapter is intended to expand and discuss the results obtained from the
analysis of the sheet pile wall at the University of South Florida, Tampa, in Chapter 4. A
total of three sheet pile wall projects will be studied: a strutted sheet pile wall in
Hotchstetten, Karlsruhe, Germany; an anchored sheet pile wall in Green Lanes, Hatfield,
United Kingdom, and two strutted sheet pile walls in Rotterdam, Sweden. Thus, four
sheet pile wall analyses will be carried out and the results will be discussed. The sheet
pile walls of the first two projects are on granular materials, whereas those in Rotterdam
are in very soft clays.
Hochstetten Sheet Pile Wall
Introduction
The Hotchstetten sheet pile wall is a full-scale sheet pile wall test performed by
the Department of Soil Mechanics at the University of Karlsruhe (Universität Karlsruhe),
Karlsruhe, Germany. The actual sheet pile wall test took place between May 25, 1993
and June 8, 1993. Before the test, a series of pretests focussed on the calibration of the
equipment were conducted both in the laboratory and in the field.
The pretests in the laboratory involved the sheet pile walls for the actual tests and
included the following:
• Test to verify the functioning of the inclinometer devices, this consists of measuringthe deflection of rectangular steel tube welded at different depths to a sheet pile wallin known inclination position.
153
• Test to check the initial condition readings (zero pressure measurements) from thepneumatic and electric earth pressure gauges, also attached to sheet pile wall KRUPPKDVI, at points marked E1 through E18 of Figure 5-1.
• Bending test, performed on sheet pile walls instrumented with strain gauges; thesestrain gauges were the devices for measuring the bending moments at the five pointsmarked with DMS or D1 through D10 in Figures 5-1 and 5-2;
• Test for calibration of the load cells for the measurements of forces developed in thestruts KRUPP Gi SV 380, Figure 5-3;
Figure 5-1. Locations of earth pressure gauges and strain gauges. (Source: Wolffersdorff,1997).
154
Figure 5-2. Location of strain gauges in bending test. (Source: Wolffersdorff, 1997).
Figure 5-3. Calibration of load cells in struts. (Source: Wolffersdorff, 1997).
The pretests in the field were necessary to check the performances of the
measuring devices under the field conditions. The following tests were included in the
field pretests:
• Verification of the functioning of the earth pressure measurement, the effect of thepile vbration during the driving process and the driving of adjacent piles wereexamined;
• Field bending test, setting up of the sheet pile wall as a cantilever beam partly driveninto the ground and loaded laterally;
• Long term tests for which the sheet pile walls had undergone different atmosphericconditions for 46 days.
155
Test Set-up
The full-scale sheet pile wall test was performed in a total land area of
10 m×10 m. The actual test wall is made of 12 sections KRUPP KD VI. The test wall
thus extends to a total length of 7.20 m. The plane strain condition was assured by
installation of two separation walls (made with bentonite) in the perpendicular direction
to the wall plane, as indicated in Figure 5-4. The wall is to be symmetrically supported
by 3 struts KRUPP Gi-SV-380, which are spaced center-to-center 2.40 m between each
other and are about 4.00 m. The depth of installation of the strut is to be at 1.25 m below
the ground surface. On the other end of the strut support, a very stiff wall made of 8
double sections of ARBED PU8 has been installed. It is assumed that the PU8 walls are
fully fixed and the lateral deflections are zero or negligible. The waling with sections
HEB 240 assured the connection between the two types of walls and three struts. On the
left and right sides of the walls, 4 double sections PU8 were installed at each side to
protect the excavated area. A water tank was also used to apply a surcharge of 10 kN/m2.
The water tank was about 7.00 m long (parallel to the wall), 4.00 m wide and 1.00 m
high. From the side view, the distance between the test wall and the closest edge of the
tank is 1.00 m. Figure 5-5 shows the soil-structure profile after the installation of the
strut and at the end of the excavation. The steel sections HEB 140 were for survey
operation purposes, which will be discussed in the next paragraphs.
The measuring devices were placed within three axes of measurements indicated
in Figure 5-6. The sheet pile inclinometer casings I11, I12, and I13 were located at the
section of the wall where the struts were installed. The earth pressure gauges and the
strain gauges (to measure the bending moments) introduced in Figure 5-1 were installed
on the sheet pile section I and section II of Figure 5-6.
156
Figure 5-4. Plan view of the sheet pile wall test. (Source: Wolffersdorff, 1997)
Figure 5-5. Side view of the sheet pile wall test. (Source: Wolffersdorff, 1997)
157
Other devices of the plan view are located farther from the test wall plane and are
not relevant to the purpose of this study. Figure 5-7 displays the different depths at
which the measuring devices are located. In addition to the gauges afore mentioned, the
targets T11 through T33 for the survey operation are shown as well.
Figure 5-6. Plan view of locations of measuring devices. (Source: Wolffersdorff, 1997)
Figure 5-7. Side view of locations of measuring devices. (Source: Wolffersdorff, 1997)
158
Actual Test and Different Stages
The actual test started with driving of the test sheet piles, followed by the first
excavation and then the installation of the struts. The last excavation was conducted in 3
excavation stages before the surcharge from the tank was applied by filling it with water.
Despite some technical problems during the process (e.g., missing contacts between the
waling and the wall, distorsion of some sheet piles due to driving, bentonite leaking from
separation wall during the excavation, circuits damaged by lightening strike, etc.), the test
was successfully terminated and useful data could be collected. The strut load cells, the
LVDTs (strain gauges) and the earth pressure gauges were connected to the data
acquisition system UPM 60 and a personal computer. The wall lateral deflections were
monitored by two methods: with inclinometer measurement system that includes the
evaluation program GLÖTZL, and the survey operation system with a Theodolite level.
Apart from the initial readings, the measurements were taken at the end of the 8 major
stages followed during the whole process, Table 5-1.
Table 5-1. Test procedure for the Hotchstetten sheet pile wall
Stage name Date Task
Stage 1 05/25/1993 Excavation down to 1.00 m
Stage 2 05/26/1993 Excavation down to 1.75 m
Stage 3 06/021993 Installation of the 3 struts
Stage 4 06/02/1993 Excavation down to 3.00 m
Stage 5 06/03/1993 Excavation down to 4.00 m
Stage 6 06/04/1993 Excavation down to 5.00 m
Stage 7 06/07/1993 Surcharge water load 10 kN/m2
Stage 8 06/08/1993 Limit state
159
06/03/1993 at 9:00a.m.:
Measurements of survey points andinclinometers
06/03/1993 at 3:00p.m.:
Excavation down to depth 4:00 m;
Installation of the 3rd level of surveymeasurement points at depth 3.00 m;
Measurements of survey points andinclinometers
06/04/1993 at 9:00a.m.:
Measurements of survey points andinclinometers
06/04/1993 at 1:00p.m.:
Excavation down to depth 5:00 m;
Measurements of survey points andinclinometers
Installation of roofing segments over theexcavation pit
06/07/1993 at 10:00a.m.:
Measurements of survey points andinclinometers
06/07/1993 at 4:00p.m.:
Filling of the water tank (water level 0.50 m);
Measurements of survey points andinclinometers
06/07/1993 at 5:00p.m.:
Filling of the water tank (water level 1.00 m);
Measurements of survey points andinclinometers
Figure 5-8. Stage 5, Stage 6, and Stage 7 for predictions. (Source: Wolffersdorff, 1997)
160
Within each stage, an appropriate schedule was followed in order to achieve the
sub-tasks, such as, prestressing the struts, installation of survey rulers, etc. The 3 stages
selected for the analyses in this study were the Stage 5, Stage 6 and Stage 7. These stages
are considered to be closest to real cases of sheet pile walls in geotechnical problems.
Details and illustration of the Stage 5, Stage 6 and Stage 8 are provided in Figure 5-8.
The variation of temperature during the test was measured. It was then realized
that the daily variation of temperature was around 6º. This was considered to be small,
regarding the possible temperature effect on the behavior of the struts. In fact, the daily
variation of the strut forces was reported as only within 15% off the average strut force.
The data from the inclinometer stations I11, I12 and I13, and the survey
measurements T11 through T33 (Figures 5-6 and 5-7) were all successfully collected.
However, only the bending moments from the sheet pile section I (Figure 5-1) could be
used because of the dysfunction of some of the strain gauges in sheet pile section II
during the actual test. The inclinometer results and the survey results agreed very well,
however the inclinometer readings on I13 were so much affected by the distortion of the
piles, and only relatively very small deflections were recorded. Thus, the inclinometer
results from the axis I13 were excluded from the data collection. All of the three strut
load cells performed well during the test.
Predictions
The data from the inclinometer stations I11 and I12 were averaged to obtain the
measured lateral deflections from the inclinometer. Similarly, the survey data from the
three stations, I11, I12 and I13 at three depths: 0.00 m, 1.25 m amd 3.00 m were averaged
for each depth to obtain the lateral deflection of the wall. As mentioned above, only the
earth pressure and bending moments from the sheet pile section I were used so no
161
averaged values were needed. Finally, the sum of the strut forces from the three load
cells was divided to the total length of the wall to obtain the strut force per linear meter
[kN/m]. The lateral deflections, the bending moments, the strut forces and the earth
pressures from each stage are presented in Appendix C for reference.
The predictions for the present study will be limited on the lateral deflections of
the wall, the bending moments, and the strut forces. As mentioned above, Stage 5, Stage
6, and Stage 7 are the stages of test selected for the predictions. The first tasks in the
predictions were the identification of material properties for calculations: the properties
of the structures: struts and sheet piles; and the properties of the soil.
Structure properties
The physical and mechanical properties of the struts KRUPP-Gi SV-380 and the
sheet piles KRUPP KD VI are given in Table 5-2 (Schanz and Bonnier, 1997).
Table 5-2. Properties of strut and sheet pile sections
Strut Sheet Pile WallProperties Units
KRUPP Gi-SV-380 KRUPP KD VI
Elastic modulus E [kPa] 210,000,000
Cross-sectional area A [cm2/m] 11.6 106.0
Moment of inertia I [cm4/m] — 968
Axial rigidity E×A [kN/m] 245,000 2,200,000
Flexural rigidity E×I [kNm2/m] — 2,033
Soil properties
An extensive soil exploration has been carried out for the sheet pile wall test at
Hochstetten. Both insitu tests and laboratory tests were performed (Wolffersdorff, 1997).
162
Among the insitu tests were the CPT and the Ménard PMT. The CPT data were used for
the analysis in this study. The data from the PMT enabled identification of the soil type
at the test site. Locations of the insitu borings are presented in Figure 5-9. The test pit
was projected to be partly inside the rectangular area bounded with dashed lines; the
vertical direction of the map being parallel to the plan of the future wall.
Figure 5-9. Locations of the CPT and PMT borings. (Source: Wolffersdorff, 1997)
Based on the PMT boring MPT1 of Figure 5-9, the soil profile consists of a
superficial mixture of brown sand and grey silt, down to depth 2.20 m, which is underlain
163
by gravelly grey silty sand, the bottom layer is from depth 3.80 m and is formed by very
sandy gravel, (Wolffersdorff, 1997).
The four CPT soundings presented in Appendix C, Figure C-1, demonstrate a
good example of reproducibility of the CPT data. Considered to produce the most
conservative soil properties, the sounding CPT D was selected for the analysis. The CPT
D boring is located on the backfill side of the wall. The maximum depth of sounding
CPT D is 10.5 m, Fg. 5-10 shows the plot of the tip resistance qc and the sleeve friction fs
versus depth. Based on this CPT data, the soil profile is rather uniform, made of mixture
of sand, silt, and gravel, down to the depth 10.5 m.
Figure 5-10. CPT D data at Hochstetten site. (Source: Wolffersdorff, 1997)
The data from CPT D sounding was further reduced and the soil properties and
soil type behavior were obtained. The properties are given in Appendix C, Table C-2.
The soil stratigraphy resulting from the soil properties is given in Tables 5-3 and 5-4
from the equivalent bow count CPT (N) data and the actual CPT data, respectively. The
soil properties were evaluated from the correlations listed in Chapter 4.
164
Table 5-3. Soil properties from boring CPT D using CPT (N)
Bottom γ a CPT (N) φ E50 Eur
(m) (kN/m3) (bl/ft) (o) (kPa) (kPa)GS = 0
2.0016.5 16 32.1 15500.0 46500.0
4.5016.5 25 34.7 19950.0 59850.0
5.5016.5 42 39.0 28500.0 85500.0
7.7519.0 26 35.0 20500.0 61500.0
10.5019.0 36 37.5 25333.3 76000.0
a Schanz and Bonnier (1997)
Table 5-4. Soil properties from boring CPT D
Bottom γ a qc φ E50 Eur
(m) (kN/m3) (kPa) (o) (kPa) (kPa)GS = 0
2.0016.5 5531.3 41.5 16593.8 49781.3
4.5016.5 10582.0 41.3 31746.0 95238.0
5.5016.5 20990.0 43.5 62970.0 188910.0
7.7519.0 11930.0 40.3 35790.0 107370.0
10.5019.0 17785.6 41.9 53356.7 160070.0
a Schanz and Bonnier (1997)
Modeling and Predictions
The modeling with the CWALSHT program was possible only for Stage 6. For
each soil-structure profile of each stage, the program searches for a realistic factor of
safety that could meet the characteristic of the current soil profile. For too high or too
low factor of safety, no solutions could be obtained. Stage 5 would be a too high factor
of safety and Stage 7 is a too low factor of safety. The results from Stage 6 were not very
promising either. CWALSHT results appeared to be unrealistic. As a reminder, the
165
CWALSHT program makes the assumption that a strutted wall is treated as a beam that
can rotate about the strut application point, but has no lateral displacement at that point.
In the finite element modeling with PLAXIS, the characteristics of the finite
element used in Chapter 4 were kept the same, 15-noded triangular element. Because of
the more complex aspect of the project, more refined meshing over the whole region was
adopted. The calculation process comprises 7 calculation stages. The 7 stages were
identically equal to the stages described in Table 5-1, omitting Stage 8. The pre-stressing
and the re-stressing of the struts to 11 kN on Stage 3 and Stage 4 were simulated as well.
Graphical presentation of the finite element simulation with PLAXIS at the end of the
excavation is given in Figure 5-11. The traction load A-A represents the surcharge
coming from the water tank. The distance between the wall and the right hand boundary
of the domain is 4.00 m because zero axial deformation of the strut is assumed to be at
the connection with ARBED PU8 (fully fixed). Both the Mohr-Coulomb model and the
Hardening Soil model were used in the analysis.
Figure 5-11. Finite element simulation with PLAXIS at Stage 7 (Unit: m).
166
Results and Discussion
The CWALSHT analysis of Stage 6 resulted in very unrealistic values of the wall
lateral deflections and the bending moments. The wall top deflection was about 33.9 mm
towards the backfill side compared to the measured top deflection 5.55 mm towards the
excavation side. The bending moment predicted with CWALSHT at depth 4.00 m was
35.53 kNm/m while the measured bending moment at the same depth was 1.56 kNm/m.
This discrepancy of values can be expected when analysing strutted or anchored sheet
pile wall with CWALSHT. As listed in Chapter 2, a number of assumptions are used in
the traditional method with CWALSHT. Unfortunately, the assumptions do not represent
more efficiently the soil-structure interaction in the sheet pile wall analysis. For instance,
the wall is treated as a beam pin supported at the strut point, which will cause no lateral
deflection of the wall at the point although the strut section does have axial elastic
properties. This assumption will affect significantly the pressure distribution applied to
the wall. Also, the assumption that the limit equilibrium state is reached in strutted or
anchored sheet pile wall analyses would generally result in a smaller magnitude of active
earth pressure and a larger passive earth pressure than those from the actual case. Based
on the results from this analysis, the CWALSHT is uneconomically very conservative as
far as the selection of sheet pile section is concerned.
In contrast to the traditional method, the finite element method with PLAXIS
results fall within reasonable ranges regarding to the measured values. The predicted
lateral deflections from the CPT (N) and CPT input parameters are plotted in Figures 5-
12 and 5-13, respectively. Unlike the traditional method, the pattern of the deformation
of the wall is closely is well predicted by the PLAXIS modeling. The accuracy of the
predictions in both types of input parameters appears to be stage-dependent: as the
167
excavation gets deeper, the numerical predictions become more conservative (greater
predicitons predicted). As it was encountered in Chapter 4, the Hardening Soil model,
and the Mohr-Coulomb model with the elastic modulus 3×Eref, generally provide better
deflection predictions. The simple Mohr-Coulomb model generally resulted in largest
deflections. For instance, at Stage 5, when the measured top deflection of the wall is
7.5 mm, the predicted top deflections from CPT (N) are 6.9 mm and 7.4 mm using
Hardening Soil model and Mohr-Coulomb 3×E, respectively. From the simple Mohr-
Coulomb model, the top deflection is 15.7 mm. This trend of the results regarding the
top deflection is also true with the CPT data, Figure 5-13.
The largest deviations of the predicted deflections from the measured deflections
are most pronounced at Stage 7 (minimum of depth of penetration 1.00 m and application
of surcharge 10 kN/m2). For the CPT (N), they occur between the depths 3.00 m and
4.00 m where for instance the maximum deflection from the Mohr-Coulomb 3×E is 9.2
mm (closest from the three different models) at depth 3.25 m versus the measured
deflection of 3.4 mm at depth 3.10 m, Figure 5-12. The possible cause of the larger
deflection discrepancy due to deeper excavations and application of the water load could
not be clearly understood. However, the results from the CPT data, Figure 5-13 showed
less increase of the deflection but still less than the increase of the measured deflections
from Stage 5 to Stage 7.
This is the direct results of the higher values of the strength and stiffness
properties of the soil derived from the CPT data, Tables 5-3 and 5-4. Thus, in this type of
calculations, the CPT data seemed to provide better deflection prediction than the blow
count based parameters.
168
STAGE 5: Excavation to 4.0m
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)
Dep
th (m
)
STAGE 6: Excavation to 5.0m
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)
Dep
th (m
)
STAGE 7: Surcharge 10kN/m 2
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)
Dep
th (m
)Figure 5-12. Measured and PLAXIS predicted lateral deflections from CPT (N).
169
Stage 5: Excavation to 4.0m
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)D
epth
(m)
Stage 6: Excavation to 5.0m
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)
Dep
th (m
)
Stage 7: Surcharge 10kN/m 2
0
1
2
3
4
5
6
0 2 4 6 8 10 12 14 16 18 20
Deflection (mm)
Dep
th (m
)Figure 5-13. Measured and PLAXIS predicted lateral deflections from CPT data.
170
Stage 5: Excavation to 4.0m
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8
Bending Moment (kNm/m)D
epth
(m)
Stage 6: Excavation to 5.0m
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8
Bending Moment (kNm/m)
Dep
th (m
)
Stage 7: Surcharge 10kN/m 2
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8
Bending Moment (kNm/m)
Dep
th (m
)Figure 5-14. Measured and PLAXIS bending moments from CPT (N).
171
Stage 5: Excavation to 4.0m
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8Bending Moment (kNm/m)
Dep
th (m
)Stage 6: Excavation to 5.0m
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8Bending Moment (kNm/m)
Dep
th (m
)
Stage 7: Surcharge 10kN/m2
0
1
2
3
4
5
6
-8 -6 -4 -2 0 2 4 6 8Bending Moment (kNm/m)
Dep
th (m
)Figure 5-15. Measured and PLAXIS bending moments from CPT data.
172
The bending moments given in Figures 5-14 and 5-15 show the same
characteristics as the deflections when compared with the measured bending moments.
At Stage 5, the least differences between the predicted and the measured bending
moments prevail. At the three stages, the Hardening Soil model and the Mohr-Coulomb
3×E model produced more accurate and more consistent results. Nevertheless, the simple
Mohr-Coulomb model from the CPT data seemed to predict conservatively well enough
the bending moments; for instance at Stage 7, the measured bending moment at depth
3.00 m was 2.76 kNm/m versus 5.85 kNm/m. Overall, the bending moment predictions
using the CPT input parameters are closer to the measured bending moments than when
using the CPT (N).
0
10
20
30
40
50
60
5 6 7Stage
Stru
t For
ce (k
N/m
)
CPT (N) input data
0
10
20
30
40
50
60
5 6 7Stage
Stru
t For
ce (k
N/m
)
CPT input data
Figure 5-16. Measured and PLAXIS strut forces.
Prediction of the strut forces with PLAXIS produced fairly good accuracy. The
Hardening Soil model and the Mohr-Coulomb 3×E model from the CPT input parameters
173
provided the best predictions with, for instance at Stage 7, 30.27 kN/m and 37.21 kN/m,
respectively compared to the measured strut force 33.72 kN/m. The worse prediction
was from using the CPT (N) input parameters with the Mohr-Coulomb model:
58.71 kN/m at Stage 7. Figure 5-16 clearly show the difference between the CPT (N) and
the CPT data; the CPT (N) was overestimating the strut forces in all of the stages except
Stage 5 with the Hardening soil model. The combination of the CPT data with the
Hardening Soil model or the Mohr-Coulomb 3×E model provided good strut forces
predictions.
Conclusions
It is realized that some of the conclusions drawn from Chapter 4 were confirmed
in the Hochstettem strutted sheet pile wall. Considering the strutting of the sheet pile
wall, the following conclusions can be stated:
The traditional method with CWALSHT cannot simulate appropriately the
behavior of a strutted sheet pile wall in Hochstetten. The assumption of the zero lateral
deflection at the strut point dramatically affects the results of analysis with the traditional
method.
The CWALSHT program does not enable analysis of case histories with factor of
safety too low or too large. In this project, the low factor of safety is described by the
soil-structure profile at Stage 7 and Stage 8 whereas the high factor of safety is described
by Stage 5 and upper (to Stage 1).
In finite element modeling with PLAXIS, the Hardening Soil model and the Mohr-
Coulomb 3×E model can give reasonably good results in the predictions of lateral
174
deflections, bending moments, and strut forces. On the other hand, the Mohr-Coulomb
model generally yields slightly more conservative results.
The combination of the Hardening Soil model or the Mohr-Coulomb model with
unload-reload modulus with CPT data gave the most consistent results in the predictions
of the lateral deflections, bending moments, and strut forces. Especially, for the present
sheet pile wall, the CPT data resulted in very good accuracy results.
Hatfield Sheet Pile Wall
Introduction
The Hatfield sheet pile wall is a part of the A1(M) tunnel improvement project at
Hatfield, Hertfordshire, United Kingdom, in 1985 and 1986; it is a temporary anchored
sheet pile wall designed by the Main Civil Contractor, Tarmac Construction, Ltd. The
designer realised that the construction of the sheet pile wall would provide a rare full-
scale test opportunity and decided to monitor the performance of the temporary anchored
sheet pile wall. The main goal was to better understand the different aspects of soil-
structure interactions: soil-sheet pile wall and soil-ground anchors. Eventually, the
project became a research work jointly executed by various contractors and Institutes. A
Steering Group was set-up, it is composed of the following: Tarmac Construction Ltd,
Heriot-Watt University, Hatfield Polytechnic, Cementation Research Ltd., Transport and
Road Reasearch Laboratory, and SERC.
The full description of the anchored sheet pile wall project was reported in a
TRRL Research Report 99, published by the Transport and Research Laboratory,
Department of of Transport, in 1987, entitled: Behavior of a Temporary Anchored Sheet
Pile Wall on A1(M) at Hatfield (Symons et al., 1987).
175
Site Description
A map of the site that includes the tunnel to be expanded and the outlying areas is
shown in Figure 5-17. The temporary anchored sheet pile wall is to be constructed at a
distance approximately 250 m north of A1(M) tunnel. The relevant section of the test
wall is shown as hached regions on the right hand side of Figure 5-17. Only 20 m out the
total research area of 250 m long 54 m wide, is instrumented for the full-scale anchored
sheet pile wall test.
Figure 5-17. Anchored sheet pile wall site at Hatfield. (Source: Symons et al., 1987)
Geology
The geological property of the site is essentially formed with Pleistocene Anglian
deposits underlain by Chalk bedrock. The predominance of superficial granular soil is
176
common in the area. Despite the possible layering variation, a lodgement till deposit
divides the sequence of gravel and sands into two parts. The depth of the till layer, based
on the SPT borings, was around 13.0 m below the ground surface.
Soil profile
An extensive site investigation carried out in 1979 in addition to a smaller
investigation in 1985 was used to establish the soil profile at the site. The 1985
investigation was achieved with four SPT borings to depths ranging from 15.5 m to 25.0
m. The locations of the SPT borings are indicated as BH1 to BH4 in Figure 5-17. Thus,
the sole insitu test performed for this project was the SPT. The SPT N values from the
borehole BH1, along with the typical soil types are presented in Figure 5-18.
Figure 5-18. Typical soil profile at Hatfield test site. (Source: Symons et al., 1987)
177
The soil profile exhibits an essentially granular soil down to the chalk layer. They are
separated by the impervious Till layer of variable thickness (Symons et al., 1987).
Characteristics of the anchored sheet pile wall
The design of the sheet pile wall was based on the limit equilibrim theory. The
depth of penetration was determinted from the stability of the soil-structure system with a
suitable factor of safety. The strength properties of the granular soils were derived from
the SPT data and using the correlations developed by Peck, Hanson, and Thornburn
(1974). The structural design for the piles and the anchors was focussed on the axial
force to be developed in the anchors, and the bending moments and shear forces in the
piles under working conditions. The effect of the change of stress state due to the
insertion of the piles in the ground was not taken into account. The method of calculation
used was the British Steel Corporation, BSC method; the following design parameters
were obtained: the depth of penetration of 3.92 m and a total length of the pile 13.0 m
(factor of safety 2.0), horizontal component of the anchor forces of 167 kN/m, and
maximum bending moments of 422 kNm/m. The pile section that meets the design
requirements was Larssen No. 3; however, the pile section Larssen 4/20 (Grade 50B),
(Appendix C, Figure C-2) was selected because of equipment compatibility and safer
stability concerns during the driving of the piles. The anchoring point is located at depth
3.2 m below the ground surface; the geometry of the anchorage system was composed of
the free length 9.0 m and the angle of inclination 25°. The fixed length (grouted part), L,
of the anchors is dependent on the soil friction angle (P = L×n×tanφ, n is an empirical
factor, see Symons et al., 1987) and was equal to 8.0 m.
178
Anchored Sheet Pile Wall Test
Instrumentation
Measuring devices were installed on the sheet piles and around the 20 m wall
length in order to monitor deformations and forces developed in the soil-structure system.
The locations of the various devices for field measurement are shown in Figures 5-19 and
5-20. Three inclinometer casings I0, I1 and I5 were attached to the sheet pile with shaped
driving shoes at their bottoms. Locations of the inclinometers I0, I1, and I5 relative to
the Larssen 4/20 sections are illustrated in Appendix C, Figure C-3. The measured
bending strains were from two piles P9 and P18 marked as “strain gauged pile” on Figure
5-19 and allowed to evaluate the actual bending moments developed in the piles. The
section modulus of the Larssen 4/20: 2266 cm3/m (British Steel Corporation, 1984).
Figure 5-19. Plan view and instrumentation locations. (Source: Symons et al., 1987)
179
During the driving, the inclinometers were protected from being damaged by
using a welded wedge. The measurement of forces developed in the anchors are
achieved by using load cells installed at points marked A1 to A7 in Figure 5-19.
Figure 5-20. Side view and instrumentation location. (Source: Symons et al., 1987)
Actual test
The actual test started with the driving of of the pile section Larssen 4/20 into the
ground, followed by the several excavation stages. At depth 3.2 m, the anchorage and
tensioning were performed; then the final excavation process took place. Table 5-5
summarizes the schedule of the test, the operation, and level of excavation at each stage.
Details on the sub-tasks performed in each stage are explained in the Research Report 99
mentioned above. The stages that were simulated in this study, are Stage 2 to Stage 5 of
Table 5-5. Stage 2 is a cantilevered type of sheet pile wall whereas Stage 3 through Stage
5 are from the prestressing of the anchor to the final excavation stage. Table 5-5 also
180
shows the relatively short period of time from Stage 3 to Stage 5. Simulation of Stage 6
will evaluate the effect of creep in the soil, the anchorage.
Table 5-5. Major construction stages and tasks
Stage name Period Major Task
Stage 1 7/25/85 – 7/31/85 Driving of the pilesStage 2 8/14/85 – 8/28/85 Excavation to 3.2mStage 3 9/10/85 – 9/20/85 Anchorage Installation and PrestressingStage 4 9/20/85 – 10/01/85 Excavation to 7.1mStage 5 10/11/85 – 11/26/85 Dewatering and Excavation to 9.3m
11/26/85 – 4/28/86 Construction of Permanent Wall4/28/86 – 5/21/86 Backfilling and Release of AnchorsStage 6 (3)5/01/86 – 7/03/86 Road Construction and Pile Extraction
Modeling and Predictions
Structure properties
The free length of the anchors was made of four prestressed strands of diameter
15.4 mm that are wrapped in denso tapes (Symons et al., 1987). The axial and flexural
properties of the anchor strands and the pile section Larssen 4/10 are given in Table 5-6.
Table 5-6. Properties of ground anchors and Larssen 4/20 50B
Anchor Cable Sheet Pile WallProperties Units
4×15.2 mm rods Larssen 4/20
Elastic Modulus E [kPa] 210,000,000
Cross Sectional Area A [cm2/m] 2.44 207
Moment of Inertia I [cm4/m] — 43,167
Axial Rigidity E×A [kN/m] 51,333 4,347,000
Flexural Rigidity E×I [kNm2/m] — 90,651
181
Soil properties
There is only one set of input parameters for the soil properties in this calculation.
The soil properties are derived from the SPT boring already presented in Figure 5-18.
The layering was then obtained by taking the average of the closer values of the SP N.
The soil properties evaluated from the correlations listed in Chapter 4 are given Table 5-
7. The layering was stopped at the impervious Till layer, which is usually considered to
be a hard layer (Bell, 1987).
Table 5-7. Soil properties from SPT N
Bottom SPT N γ φ E Eur
(m) (bl/ft) (kN/m3) (°) (kPa) (kPa)GS = 0
3.0 74.0 23 44.6 44500.0 133500
5.0 11.5 18 30.6 13250.0 39750
7.5 14.7 19 31.6 14833.3 44500
13.0 41.7 21 38.9 28333.3 85000
GS = Elev. +77.75mGWT = Elev. +68.50m
Modeling
The modeling with CWALSHT was possible for Stage 4 and Stage 5 in this
project. Calculations for Stage 2 and 3 could not be performed with CWALSHT
probably because of too high factor of safety in the soil-structure stability. Deflections
and bending moments of Stage 4 were obtained from the design mode of calculation.
The important requirement in the design mode was that the depth of penetration comes
out approximately equal to the actual depth of embedment at Stage 4. The inclined
ground anchorage is modeled with a horizontal strut support type, which can be accepted
as reasonable by neglecting the effect of the downward component of the anchor forces to
182
the soil-structure behavior. However, the prestressing of the anchorage cannot be
modeled as the anchor point will be assumed as only a pin support on the sheet piles.
In modeling with PLAXIS, the type of modeling similar to that of the Karlsruhe
strutted wall was used. However, a few differences are emphasized: for the anchorage
system, the grout is modeled with geotextile finite elements whose properties were
discussed in Chapter 2, and the free length is modeled with tie-rods. Geometrically, the
distance between the sheet pile wall and the right hand side boundary was set to 30 m.
Figure 5-21. Modeling with PLAXIS of Hatfield sheet pile wall at Stage 5 (Unit: m).
This was based on the zero lateral displacements and settlements of soil measured at
distances of about 20 m from the wall (LL and S Stations of Figure 5-19). Figure 5-21
displays the soil, the sheet pile, the anchor free length (tie rods), and the geotextile finite
elements.
Results and Discussion
The predicted lateral deflections and the measured deflections from the three
inclinometer casings I0, I1, and I5 are compared and shown in Figure 5-22. The finite
183
element results were plotted assuming the base fixities. The lateral deflections in the
cantilevered wall of Stage 2 were well predicted by all of the constitutive models. The
simple Mohr-Coulomb model produced a lateral deflection of the wall bottom larger than
the top deflection, 1.51 mm and 1.35 mm, respectively. This is considered to be the
result of heaving on the excavation side of the wall. Such magnitude of lateral deflection
due to heaving from an excavation to a depth 3.2 m would not be very realistic. The
excessive heaving is common with the Mohr-Coulomb model. The Hardening Soil
model best simulated the deformed line of the wall; the difference becomes larger as one
goes up to the top of the wall: the Hardening Soil model followed a straight line whereas
the measured deflections are slightly concave curves into the backfill. In the theory of
cantilevered beam in flexure, the deformed line from the Hardening Soil model would be
more realistic. In Stage 3, when the ground anchors were installed and then prestressed;
the lateral deflections become negative, and it is most pronounced with the Mohr-
Coulomb model. The Hardening Soil model and the Mohr-Coulomb 3×E model were
very similar and best predicted the wall deformation. The measured top deflection from
Station I0 was -2.2 mm versus -3.3 mm and -1.1 mm from the Hardening Soil model and
the Mohr-Coulomb 3×E model, respectively. At Stage 3, the actual deflected wall shows
that no lateral movement of the soil occurred from Elevation +73.25 m down. This was
not obtained from the numerical modeling. At Stage 4, the discrepancies appear more
evident. The results from CWALSHT were significantly different from the measured
values, the reason of this was already discussed in the case of Karlsruhe wall.
Unexpectedly, the deflections in Stage 5 were generally well predicted with CWALSHT,
it is considered to be the ideal soil-structure profile that best fits the calculation by
184
CWALSHT. For the constitutive models, at Stage 5, the Hardening Soil model becomes
more conservative, this is probably due to considerable decrease of the stiffness in larger
deformations.
The bending moments measured from the piles P9 and P18 (Strain gauged piles of
Figure 5-19) were compared with CWALSHT and the finite element bending moments in
Figure 5-23. Generally, all of the constitutive models overpredicted the bending
moments in the four stages. CWALSHT analysis at Stage 4 and Stage 5 gave a good
estimation of the bending moment at the anchor level; however, the maximum bending
moments that occur around the mid-depth of the wall were unconservatively
overpredicted. The finite element constitutive models produced more consistent
predictions, except at Stage 5 where the Hardening Soil model significantly overpredicted
the peak moments in the wall. It is worth mentioning that reliance on the magnitudes of
the measured bending moments is not assured because of the damage caused by
installation of the waling, (Symons et al., 1987). In fact, some of the strain gauge
readings at the anchor points and above the anchors were lost after installation of the
waling. The accuracy of the bending moments were estimated to be ±15 kN/m. At Stage
3, the measured bending moment at the anchor level was around 25 kNm/m, but the
predicted values (not included in the plots) were 106.3 kNm/m, 82.7 kNm/m, and 70.5
kNm/m for simple Mohr-Coulomb, Hardening Soil, and Mohr-Coulomb 3×E,
respectively. These values are unrealistically high. In their finite element modeling, Day
and Potts (1991) also encountered this large difference. It was explained that the finite
element analysis considers the anchor force to be a point load rather than distributed one
over a finite length.
185
Stage 2: Excavation to 3.2m
64
66
68
70
72
74
76
78-15-10-50510152025
Deflection (mm)
Elev
atio
n (m
)
Stage 3: Anchor and Prestress
64
66
68
70
72
74
76
78-15-10-50510152025
Deflection (mm)
Elev
atio
n (m
)
Stage 4: Excavation to 7.1m
64
66
68
70
72
74
76
78-15-10-50510152025
Deflection (mm)
Elev
atio
n (m
)
Stage 5: Excavation to 9.3m
64
66
68
70
72
74
76
78-15-10-50510152025
Deflection (mm)
Elev
atio
n (m
)
Figure 5-22. Lateral deflections of Hatfield sheet pile wall.
Stage 2: Excavation to 3.2m
64
66
68
70
72
74
76
78-75-50-250255075100125150
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 3: Anchor and Prestress
64
66
68
70
72
74
76
78
-150
-125
-100
-75
-50
-250255075
100
125
150
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 4: Excavation to 7.1m
64
66
68
70
72
74
76
78-75-50-250255075100125150
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 5: Excavation to 9.3m
64
66
68
70
72
74
76
78-75-50-250255075100125150
Bending Moment (kNm/m)
Elev
atio
n (m
)
Figure 5-23. Bending moments of Hatfield sheet pile wall.
In the actual anchored sheet pile wall, the anchor forces are distributed evenly by
the waling. As a result of the large bending moments at the anchor level, the depth of the
maximum bending moments on the other side of the piles moves to a lower position. At
Stage 2, good agreements could be obtained. From Stage 3 on, the good agreements
occurred only above the anchor level; below the anchor level, the most accurate bending
moment prediction was from the Mohr-Coulomb 3×E model. For instance at Stage 5, the
maximum measured bending moment is around 65.0 kNm/m at Elevation +71.75 m,
whereas the corresponding value from the Morh-Coulomb 3×E was 80.2 kNm/m. The
discrepancy between the values of the bending moments was aggravated by the
overprediction of moment values at the anchor level.
Comparison of the measured anchor forces with the predicted values is presented
in Table 5-8. The force magnitudes represent the horizontal components of axial forces
developed in the strands per unit length. The CWALSHT program produced inconsistent
results in estimation of the anchor forces. This is not unexpected because the two stages
were calculated with different modes (design mode and analysis mode). The factor of
safety value that was input for the design mode (Stage 4) was 2.5, compared to the factor
of safety 1.0 in the analysis of Stage 5. The use of factor of safety 2.5 was the only
doable calculation that generates the depth of penetration corresponding to that of Stage
4: 5.9 m. Unlike the CWALSHT, the finite element model with PLAXIS showed more
consistent results; the results are generally larger than the measured forces but the
increase in anchor forces due to deeper excavation was duplicated. The lower values of
the measured anchor forces can be attributed to the creep effect in the strands.
188
Table 5-8. Horizontal components of anchor forces
Measured CWALSHT PLAXIS MC PLAXIS MC 3×E PLAXIS HSStageUnits [kN/m]
Stage 4 153 183.4 185.0 174.1 182.9
Stage 5 157 96.3 192.4 178.1 198.9
A part of the axial forces have been released. In the finite element analysis, the strands
behave purely elastically. However, the compared values are reasonably good for
analysis and design purposes. Again, the Mohr-Coulomb 3×E model provided the most
accurate estimation of the anchor forces.
Conclusions
Additional useful conclusions could be drawn from the Hatfield anchored sheet
pile wall project. It was realized that the anchored sheet pile wall behaves differently
compared with cantilevered sheet pile walls and strutted sheet pile walls. Rather than just
an unloading case, the prestressing of the ground anchor causes lateral compression of
soil around the anchor point. The anchor force, when modeled with the finite element
method generates locally very high values of bending moments and changes the moment
distribution along the sheet pile sections. Simulation of the behavior of the anchor
strands could not be achieved satisfactorily because of the creep behavior in the strands.
Nevertheless, the results showed that finite element code PLAXIS can be a useful tool for
analysing the behavior anchored sheet pile walls. The following conclusions can be
stated:
The SPT derived soil properties did not yield good accuracy and good consistency
results in analyzing the Hatfield anchored sheet pile wall. The procedure followed when
189
performing the SPT, and the energy efficiency associated to the equipment were not
known. An evident sign of unreliability of the SPT data is the too high blow count N =
74 for the granular fill down to depth 3.00 m. The other factor of the result discrepancies
could be the malfunctioning of the measurement devices (bending moments), as
explained by Day and Potts (1991).
The traditional method CWALSHT has a very limited capacity in analysing
anchored sheet pile walls. This applies in the geometry of the soil-strucuture profile as
well as in the modeling of the prestressing of the anchors.
The finite element modeling with PLAXIS can be used as a conservative method in
analysing the behavior of anchored sheet pile walls; this can be applied in the sizing of
the pile from the maximal bending moments, the sizing of the cross sectional area of the
anchor rods (or strands) and the lateral deflections of the wall.
The three constitutive models used in this analysis can be reliable within
reasonable margins in predicting the lateral deflections, bending moments, and the anchor
rod forces. The most accurate constitutive model was the Mohr-Coulomb 3×E model,
followed by the Hardening Soil model, and finally the simple Mohr-Coulomb model.
Rotterdam Strutted Sheet Pile Walls in Very Soft Clay
Introduction
The Rotterdam strutted sheet pile walls are a full-scale sheet pile wall test
performed by the Geotechnical Laboratory of the Delft University of Technology, in
Pernis, a suburb of Rotterdam, the Netherlands between April 1999 and January 2000.
The main purpose of the full-scale test was to examine the performance of the steel piles
regarding the recent development of the European Standards such as Eurocode 3 (Steel
structures) and Eurocode 7 (Geotechnical design). The sheet pile wall test in a very soft
190
clay was also the object of research on the performance of sheet piles with plastic hinges,
with oblique bendings (double U-sections), and in short-term and long-term conditions.
The test became a large project involving a number of subcontractors (monitoring and
instrumentation; sheet pile provider; insitu and laboratory tests, etc.) and administration
tasks, which were under the supervision of a specially created international scientific
committee (Kort, 2002). The Rotterdam sheet pile wall test site covered a land area of
20m×50m; two walls were to be tested and they are made of two different sheet pile
sections: AZ13 sections and U-piles Larssen 607K sections. Pretests were carried out on
the sheet pile section AZ13 in order to check the material properties and verify the
procedure of measurement and the functioning of the measuring devices (strain gauges,
bending moments). The pretests were primarily composed of bending tests and tensile
tests in the field. Details on the pretests can be found in the work by Kort (2002).
Test Site and Soil Profile
Geology
The site is mainly formed by superficial soft clay and peat layers of about 16.5 m
thick, which are underlain by a Pleistocene sand layer. The sand layer is connected to a
river called Maas found about 2 km from the site. The ground water table is about 1.0 m
below the ground surface.
Soil profile
A number of laboratory and insitu tests were performed for the sheet pile wall
project in Rotterdam. Among the insitu tests performed were CPT, CPTU test
(piezocone test), Vane shear test (V), Ménard PMT (MPM) and sampling (B). The
locations of the insitu borings relative to the sheet pile walls are found in Kort (2002).
The four CPT and the two CPTU tests were stopped at depth 20.0 m. The boring logs
191
from the CPTU displays the level of the water table from the initial pore pressure
measurements. Kort (2002) also included the boring log of sounding CPT2, the data
which were used for analysis in this study. The other five boring logs are presented in
Appendix C. The sounding CPT shows very low values of the tip resistance (qc <
1.0MPa) with rather high values of the friction ratio (Rf = 10%). This indicates the very
soft clay layers or peat down to the depth 17.0 m, from where the sand layer starts with
much higher value of the tip resistance and lower friction ratio.
Test Setup
The sheet pile wall test was to be performed on the walls of sections: 10 ARBED
AZ13 and 7 Larssen 607K, on the north and south sides; the two other walls on the east
and west sides are made of single U-piles LX32 (British). The length of the test piles is
19.0 m while the LX32 piles are 21.0 m. Inclinometer casings were welded onto the
north test piles A1, A4 and A5; 40 strain gauges were installed on pile A4. For the piles
Larssen 607K, inclinometer casings were welded to piles H2, H4 and H6; 40 strain
gauges were installed on the pile H5. The two separation walls on each side of the test
walls were installed in order to achieve plane strain behavior. The separation walls are
made of bentonite-soil columns of diameter 50 to 70 cm and the depths of their bottoms
range from depths 13.5 m for the closest to the test wall to 3.5 m for the farthest. The
side view of the soil-structure system (Kort, 2002) shows that the top of the test walls is
at Elevation +1.0 m and the bottom Elevation is at –18.0 m (NAP: ‘Elevation’ for the
Netherlands). The struts were made with 2 HE 300A and were installed at Elevation
+0.75 m (center).
The connections between the struts and the walings, made with HE 400B were
assured with hinged connections. The forces in the struts were transmitted into a
192
perpencidular stiff beam HE 600B in order to make the strut forces from the two test
walls act independently.
Extensive instrumentation and monitoring of the test area was conducted. The
measurements of interest in this research are the inclinometer measurements, the strain
gauges for the bending moments, and the load cells for the strut forces. The locations of
the inclinometers and the strain gauges were already mentioned above. For the forces in
the struts, 4 load cells were used to measure the forces developed in the HE 300A struts.
Test Procedure
The test procedure was composed of 6 major stages with a total of 20 subtasks
throughout the whole test. Only the first stage corresponds to the short-term analysis of
the wall behavior. Stage 2 and higher were aimed at studying the long-term behavior of
the sheet pile walls in soft clays. The schedule followed during the execution of Stage 1
through Stage 6 is presented in Table 5-9. It is worth mentioning that the maximum
depth of excavation reached throughout the whole test (up to Stage 9) was Elevation –7.0
m, Table 5-9. The rest of the stages consisted mainly of adding the sand fill on the AZ13
wall side, raising and lowering the water level in the excavated side.
Table 5-9. Sub-tasks and schedule in short-term field test
Stage name Period Sub-task
Stage 1.1 4/13/99 – 4/14/99 Dry Excavation to –4.0 m
Stage 1.3 4/19/99 – 4/20/99 Excavation under Water to –7.0 m
Stage 1.9 5/11/99 – 5/11/99 Lowering Water Level to –5.0 m
Stage 2.4 5/21/99 – 5/21/99 Lowering Water Level to –3.5 m
Stage 5.1 10/05/99 – 10/05/99 Lowering Water Level to –6.0 m
193
Five sub-stages out of 20 were selected for the analysis with CWALSHT and
PLAXIS. They were considered to be most representative of the variety of soil-structure
profiles throughout the whole test. The selected substages are Stage 1.1, Stage 1.3, and
Stage 1.9, Stage 2.4, and Stage 5.1, Table 5-9.
Modeling and Predictions
Structure properties
Both test walls were analyzed in this study to examine the repeatability of the
results as it is the only sheet pile wall project in soft clays so far. The properties of the
struts HE 300A, the sheet piles AZ13 and Larssen 607K are presented in Table 5-10. The
equivalent length for the struts is 2.5m: distance between the wall and the perpendicular
stiff struts HE 600B where the compression of the wall struts is assumed to be zero.
Drawings of the cross sections of the sheet piles AZ13 and Larssen 607K and the relative
positions of the inclinometer casings are provided in Appendix C.
Table 5-10. Properties of struts, AZ13 wall and L607K
Strut Sheet Pile WallProperties Units
2×HE 300B AZ13 L607K
Elastic modulus E [kPa] 210,000,000
Cross-sectional area A [cm2/m] 46.2 137.0 244.0
Moment of inertia I [cm4/m] — 19,700 70,030
Axial rigidity E×A [kN/m] 970,200 2,877,000 5,124,000
Flexural rigidity E×I [kNm2/m] — 41,370 147,063
194
Soil properties
Two sets of input parameters for the soil properties were obtained from the data
reduction of the CPT2 sounding: the CPT (N) and the CPT properties. The data reduction
with the program ‘Cptintr1’ is given in Appendix C. The soil stratigraphy and the soil
properties for input parameters in CWALSHT and PLAXIS are given in Tables 5-11 and
5-12, for CPT (N) and CPT data, respectively
Table 5-11. Soil properties from CPT (N) data
Bottom γ a CPT (N) φ su E(m) (kN/m3) (bl/ft) (°) (kPa) (kPa)
GS = 0.651.60 18.0 — 35.0 — 45000.0
5.25 16.6 5.9 — 103.67 18141.7
9.00 10.3 5.0 — 92.40 16169.4
15.00 15.1 4.3 — 82.34 14410.4
16.00 14.3 4.0 — 78.68 13769.5
17.50 16.2 5.5 — 98.96 17318.0
EOB=20.00 20.0 17.1 32.4 — 16050.0a Kort (2002)
Table 5-12. Soil properties from CPT data
Bottom γ a qc φ su E(m) (kN/m3) (kPa) (°) (kPa) (kPa)
GS = 0.651.60 18.0 - 35.0 - 45000.0
5.25 16.6 690.6 - 62.87 3798.4
9.00 10.3 439.3 - 32.75 2416.1
15.00 15.1 426.7 - 25.74 2346.7
16.00 14.3 415.0 - 19.12 2282.5
17.50 16.2 610.0 - 36.83 3355.0
EOB=20.00 20.0 5521.0 36.0 - 16563.0a Kort (2002)
195
.The properties of the sand fill were derived from the peak value of the (σ - ε )
diagram from triaxial compression test (Kort, 2002). Compared to Eq. (3.41), Chapter 3,
the correlation between the undrained shear strength and the blow count N, Eq. (3.25)
appears to over-estimate the su values.
Modeling with CWALSHT and PLAXIS
Because the two walls were analysed together in the same model, the soil-
structure profile is different from the typical ones. With CWALSHT, a few problems
were encountered. It is not possible to model two walls on the same profile, and
modeling in very soft clay does not suit the type of analysis in the traditional method.
None of the soil-strucuture profiles could be analysed with CWALSHT. Attempts to
solve the problem with Mathcad spreadsheets using the free earth method lead to the
conclusion that the passive forces coming from the embedded part of the pile are larger in
magnitude than the active forces from the backfill side. Thus, the equilibrium of force
would result in a pulling (rather than pushing) of the wall by the struts. The assumption
made here was undrained analysis.
The problem is handled differently in the finite element analysis with PLAXIS.
The two walls could be included in one model and simulation of the different scenarios is
achievable: seepage problem, excavation stages, etc. As for the constitutive models,
modeling the soft clays with the Hardening Soil model did not cause any major problems;
the strength parameters used were the undrained parameters. This applicability of the
Hardening Soil model to soft layers have been supported by PLAXIS users such as,
Freiseder and Schweiger (1999), and Callisto et al. (1999) (Kort, 2002). The exception is
that a too low oedometer modulus (same order of the reference modulus) had to be re-
196
evaluated in PLAXIS. In this study, the proportion between the reference modulus refE50
and the odometer modulus refoedE has been established as (using undrained conditions:
φ = 0 analysis)
( ) refrefoed EE 503.0 to5.2 ×= (5.1)
The interface property chosen between the steel and the clay layers was 0.5. The plane
strain finite element modeling of the whole soil-structure is shown in Figure 5-24 (Stage
1.9). The wall on the right hand side represents the AZ13 sections and that of the left
hand side the L607K. The calculation process was performed in five stages that simulate
Stages 1.1 through Stage 5.1 under Plastic type of calculation.
Figure 5-24. Finite element modeling with PLAXIS at Stage 1.9 (Unit: m).
197
Results and Discussion
The predicted deflections of the sheet pile walls AZ13 and L607K are compared
with the measured ones in Figures 5-25, 5-26, and 5-27, respectively. Clearly, the input
data from the CPT (N) where the undrained shear strength su is evaluated directly from
the CPT equivalent blow count did not produce good deflection predictions. The
deflections were significantly underestimated. This is attributed to the high values of the
undrained cohesion su and the modulus E, this latter was directly estimated from su,
(Bowles, 1996). On the contrary, the soil properties estimated from the CPT data
provided good predictions of the deflections. This was encountered for both walls. In
the CPT data, the modulus E was evaluated from the tip resistance qc, Eq. (3.43). It
appears that the cohesion values, Eq. (3.41) give good results when combined with the
correlation of Eq. (3.43).
For the constitutive models, generally the Hardening Soil model predicted with
consistency and better accuracy the deflections over the simple Mohr-Coulomb model
and the Mohr-Coulomb 3×E model. Unlike what has been encountered so far, the Mohr-
Coulomb 3×E model gave poor results; this could be due to the assumed elastic behavior
with high modulus. Assumption of the soft clay behaving as an elastic material would
not be adequate. This could also be an explanation of the good results from the
Hardening model, where the plastic behavior is simulated by the stress dependent
modulus of the soft layer. The last 3 stages in Figures 5-26 and 5-27 show the large pile
deformations from the Hardening Soil model
The results of the predictions of the bending moments were slightly different in
the walls AZ13 and L607K, as they are shown in Figures 5-28, 5-29, and 5-30,
198
respectively. While the Hardening Soil model results were the closest to the measured
bending moments in the wall AZ13, the Mohr-Coulomb 3×E model appeared more
accurate in the wall L607K. The explanation for the difference could be the large
flexural rigidity of the section L607K. The bending moments caused by the smaller
lateral deflections can be as high as the actual bending moments in the stiff beam. At
early stages, the bending moments from the simple Mohr-Coulomb model were generally
larger than the actual values but at larger moments (larger deformations) the actual values
become larger which can be explained by the plastic behavior of the soil, not represented
in the Mohr-Coulomb model.
The forces developed in the struts were compared for each wall in Figure 5-31.
Due to the poor results (lateral deflections and bending moments) from the CPT (N) data,
only the results from using the CPT data were shown in Figure 5-31. As encountered in
the two preceding walls, the finite element modeling with PLAXIS results in more
conservative values of the strut forces. The Hardening Soil model and the Mohr-
Coulomb model follow closely the increment paths of the force during the different
stages. The strut forces in wall L607K from the three constitutive models were larger
than those in wall AZ13, which is justified because of the larger flexural and axial
rigidities of wall L607K. The addition of the sand mound is reflected in the last stage
(Stage 5.1) when the strut forces from AZ13 become greater than those from L607K.
Within the same wall, the larger the stiffness of the soil the smaller the strut forces
developed, which could explain the relative variation of the forces during the process.
199
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)El
evat
ion
(m)
Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Stage 1.9: Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Figure 5-25. Lateral deflections of AZ13 wall using CPT (N)
200
Stage 2.4: Sand Mound
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Stage 5.1: Water Level to -6.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Figure 5-25. continued.
201
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)El
evat
ion
(m)
Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Stage 1.9: Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Figure 5-26. Lateral deflections of AZ13 wall using CPT data
202
Stage 2.4: Sand Mound
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100 200
Deflection (mm)
Elev
atio
n (m
)
Stage 5.1: Water Level to -6.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Deflection (mm)
Elev
atio
n (m
)
Figure 5-26. continued.
203
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-150 -100 -50 0 50
Deflection (mm)
Elev
atio
n (m
)Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-150 -100 -50 0 50
Deflection (mm)
Elev
atio
n (m
)
Stage 1.9: Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-150 -100 -50 0 50
Deflection (mm)
Elev
atio
n (m
)Figure 5-27. Lateral deflections of L607K wall using CPT data.
204
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)El
evat
ion
(m)
Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 1.9: Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Figure 5-28. Trend of bending moments of AZ13 wall using CPT (N).
205
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)El
evat
ion
(m)
Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 1.9:Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Figure 5-29. Bending moments of AZ13 wall using CPT data
206
Stage 2.4: Sand Mound
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 5.1: Water Level to -6.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Figure 5-29. continued.
207
Stage 1.1: Excavation to -4.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)El
evat
ion
(m)
Stage 1.3: Excavation to -7.1m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Stage 1.9: Water Level to -5.0m
-18
-16
-14
-12
-10
-8
-6
-4
-2
0
-400 -300 -200 -100 0 100
Bending Moment (kNm/m)
Elev
atio
n (m
)
Figure 5-30. Bending moments of L607K wall using CPT data.
208
AZ13 Strut Forces
0
25
50
75
100
125
150
175
200
1.1 1.3 1.9 2.4 5.1
Stage
Stru
t For
ce (k
N/m
)
L607K Strut Forces
0
25
50
75
100
125
150
175
200
1.1 1.3 1.9 2.4 5.1
Stage
Stru
t For
ce (k
N/m
)
Figure 5-31. Strut forces from AZ13 and L607K.
Conclusions
The analysis of two sheet pile walls on very soft clays provided further
understanding of the applicability of the traditional methods and the finite element
method to sheet pile wall problems. The following conclusions were obtained from
studying the Rotterdam full-scale test on sheet pile walls:
The sheet pile walls in very soft clay could not be solved by the traditional
method with CWALSHT. The deep penetration of the wall lead to the conclusiont that
strutting was not necessary, which was in contrast of what was observed in the real case.
The finite element with Πλαξισ is a useful tool to model and analyze the
behavior of sheet pile walls in very soft clays even in complicated soil-structure profiles.
The CPT can be a reliable test in analyzing sheet pile walls in soft clays; the
correlation for estimating the undrained cohesion based on the bearing capacity theory
provides a good estimate of the su values. The equivalent SPT blow count CPT (N), or
implicitly the SPT N, did not give satisfactory results. The direct relationship between
the SPT N and the undrained cohesion su is not recommended.
209
The Hardening Soil model in PLAXIS can be used to model the behavior of very
soft clays in sheet pile wall analyses. The undrained properties could be used for the
model efficiently without inputting the effective strength parameters c’ and φ’. The
Hardening Soil model does not possess any creep parameters however the creep behavior
of the clay was simulated in the stress dependent unload-reload modulus of the soil.
Conclusion for Sheet Pile Walls
The four sheet pile walls in the three different projects have been studied. This
chapter was distinguished by the diversity of the wall type: cantilevered, strutted, and
anchored sheet piles with different pile sections; and the soil type, granular soils, and
very soft clay. The insitu test data, unfortunately, were only from the SPT and the CPT.
The Push-in PMT was also performed in Moffitt Cancer Center wall. Data from the
DMT and from any other type of PMT were not available. Nevertheless, very useful
findings were achieved from the study of Moffit Cancer Center cantilevered wall to the
Rotterdam strutted sheet pile walls.
Firstly, the SPT proved inconsistent. The CPT is more consistent and the
equivalent blow count CPT (N) can be equally reliable. The test procedure and
equipment are factors that can explain the difference between the two tests. The SPT in
Florida U.S.A and in Hatfield, United Kingdom were very likely to be performed under
different standards. The results are probably affected by the difference in the type of
equipment, and energy efficiency. Yet the same correlations were used to derive the soil
properties. The CPT (electric) was more robust. Although several countries use the
CPT, the equipment and the procedures do not vary as much. All use the 60° apex angle,
10 cm2 cone projected area, and the rate of penetration 20 ± 5 mm/s required by the
210
IRTP1 and the ISSMFE2 (Lunne et al., 1997). The CPT provides two sets of data, the
equivalent blow count CPT (N) and the CPT data. They both were useful and reliable.
Clearly, the CPT is recommended over the SPT and the Push-in test for the sheet pile
walls. For SPT and PMT, the standard test procedures have to be followed (e.g., for SPT,
ASTM D 1586 in U.S.A, etc.).
The traditional method with CWALSHT is not recommended to analyze
unloading cases in sheet pile walls. CWALSHT was helpless for both sheet pile walls in
sand and sheet pile walls in soft clays. The finite element method with PLAXIS has much
better capability and versatility in simulating the behavior of the walls. Any of the
constitutive models: Mohr-Coulomb, Hardening Soil, and unload-reload Mohr-Coulomb
are reliably usable. The finite element method also deals with the different type of insitu
test efficiently, the SPT, PMT, and the CPT.
1 International Reference Test Procedure2 International Society of Soil Mechanics and Foundation Engineering
211
CHAPTER 6CIRCULAR FOOTING AT GREEN COVE SPRINGS
Introduction
Assessment of footing settlement is critical in assuring the serviceability criterion
of the structure that it supports. This chapter deals with second full scale test that the
University of Florida performed for the research project financed by the Florida
Department of Transportation (FDOT). It consists of the prediction of the settlement of a
circular shallow concrete footing in Green Cove Springs located at the Headquarters of
the Applied Foundation Testing, Inc. (AFT, Inc.) and the Pile Equipment, Inc., Clay
County, Florida, Figure 6-1.
Figure 6-1. Florida state and shallow footing project entities.
212
The University of Florida conducted a static load test on a circular shallow
concrete footing in collaboration with Applied Foundation Testing, Inc.
Objectives
The main objective is to predict the settlement due to static loading using the
traditional methods discussed in Chapter 2, and the finite element method with PLAXIS;
then compare the predictions with the measured settlement in the field. The calculations
are based on the soil properties obtained from the insitu tests conducted by the University
of Florida. The insitu tests are: Electric Cone Penetration Test (CPT), the Dilatometer
Test (DMT) and the Push-in Pencel Pressuremeter Test (PMT). In addition to these tests,
the LawGibb Group also performed the Standard Penetration Test (SPT) and soil
sampling.
Scope of the Work
The following tasks included in this chapter to attain the objectives above are:
• Presentation of the data from SPT, CPT, DMT, and PMT and identifying the inputparameters for the theoretical predictions.
• Construction of the circular concrete footing and determination of its physical andmechanical properties.
• Instrumentation of the footing with reference beams, load cells, and LVDTs.
• Loading of the footing with weights, data acquisition during the loading process, andmeasurement of the actual settlement of the concrete footing.
• Comparison of the LVDT measured settlement with the results obtained from back-up survey measurements.
• Predictions of the settlement using the CSANDSET for the conventional methods andthe PLAXIS code for the finite element analysis.
• Comparison of the LVDT measured settlement with the results from the theoreticalcalculations.
213
Site Description and Insitu Testing
Site Geological Characteistics
Clay County is formed by Pleistocene and Pliocene outcrop and shallow subcrop
rocks and is a potential producer of sand and gravel, clay, peat, and heavy minerals in the
state of Florida. This region of Green Cove Springs has a heavy mineral sand of
Pleistocene age, combined with fine-grained sediment deposit (Randazzo and Jones,
1997).
Insitu Test Locations
The soundings for the insitu testing were located at a distance about 10.0 ft
(maximum) from the center of the projected footing as illustrated in Figure 6-2. The
concrete footing was cast in an area confined between a quonset hut and a hardware
house. The minimum distance between the edge of the footing and the house on the
north side and on the south side are approximately 4.0 ft and 6.0 ft. The access to the
testing site can be from the southwest side (fenced) of the footing or the northeast side
(free). Figure 6-2 indicates that the total working area at the footing site was about 450
ft2 of land.
The University of Florida cone truck was used for the CPT, DMT, and PMT at 6
different depths. The CPT was conducted down to depth 34.4 ft (10.5 m) below the
ground surface; the DMT was performed to depth 27.6 ft (8.4 m); and the PMT were at
depths 3.3 ft (1.0 m), 6.6 ft (2.0 m), 9.8 ft (3.0 m), 16.4 ft (5.0 m), 23.0 ft (7.0 m), and
32.8 ft (10.0 m).
Insitu Test Data
The SPT N values versus depth from the test performed by LawGibb Group are
plotted in Figure 6-3. Samples were recovered at the depths of the measured blow
214
counts; and the description of the stratigraphy could be obtained. The soil profile is
essentially brown to grey (or mixture) silty sands down to the depth 20.0 ft (6.0 m) below
the ground surface, the sand layers are then underlain by a very soft grey clay layer.
Attempt to recover intact samples were usuccessful because of the shallow depth of the
groundwater table; the SPT boring enabled to estimate that the ground water table at the
time of the test was 5.5 ft (1.68 m) below the surface.
Figure 6-2. Area of load test and locations of insitu test borings.
Projected FootingProjected 6.0ft diameter Footing
CPT
DMT
PMT
SPT
15.0ft
30.0
ft
Adj
acen
t Bui
ldin
g
Adj
acen
t Bui
ldin
g
Fence
Access Area
Samples forOedometer Test
AFT, Inc., Green Cove Springs main office Building
~40f
t
215
The SPT N values indicate that the subsurface layer is composed of medium
dense sand overlaying a very soft clay at depth 25.0 ft (7.5 m). The SPT equipment used
was an automatic hammer with sampler Shelby tube of 3.0 in. diameter, and the energy
efficient was not known. Attempt to recover intact samples were usuccessful because of
the shallow depth of the groundwater table; the SPT boring enabled to estimate that the
ground water table at the time of the test was 5.5 ft (1.68 m) below the surface. The SPT
N values indicate that the subsurface layer is composed of medium dense sand overlaying
a very soft clay at depth 25.0 ft (7.5 m). The SPT equipment used was an automatic
hammer with sampler Shelby tube of 3.0 in. diameter, and the energy efficient was not
known.
The data resulting from CPT are presented in Figure 6-4, the soil properties after
the data reduction with Cptintr1 are given in Appendix D, Table D-1. The CPT sounding
also indicates an overconsolidated sandy layer in the subsurface, suggested by a high
value of both the tip resistance and the friction sleeve (14MPa and 77kPa, respectively);
and a very soft clay with extremely low values of the tip resistance and the friction
sleeve.
The data from the DMT are plotted in Figure 6-5, the water table level estimated
from the SPT is added to the plots. An examination of the horizontal stress index KD
versus depth also suggests that the overconsolidated sand layer in the soil profile very
likely. Based on the DMT data, the bottom of that layer is probably around 9.0 ft (3.0 m)
below the ground level. The soil properties resulting from the data reduction with the
‘Dilatometer’ program are given in the Appendix D, Table D-2.
216
Loose Grey Silty Sand
Very Loose Grey Silt
Very Soft Grey Sandy Silt
Very Soft Grey Silty Clay
Brown Silty Sand
Grey Silty Sand
Grey/Brown Silty Sand
Loose Grey Sandy Silt
13
14
15
10
6
2
2
3
0
1
2
3
4
5
6
7
8
0 5 10 15 20
SPT N (blows/ft) D
epth
(m)
Figures 6-6 and 6-7 show the corrected curves of the PMT pressure versus
volume. The stiff sand layer is identified by the steep slope of the primary loading at
depths 1.0m and 2.0m; the clay layer at the bottom is also identified by the flat curves
(almost horizontal) which is a form of creep, in the last two depths 7.0 m and 10.0 m.
The effect of disturbance due to the friction reducer was most noticeable at depth 5.0 m.
When performing the insitu tests, a hard thin layer (crust) of about 0.5 ft thick
was encountered, therefore, a pre-bored hole was made before the CPT cone and the
DMT blade were inserted into the ground.
Figure 6-3. SPT sounding at Green Cove Springs.
217
Figure 6-4. CPT sounding at Green Cove Springs.
218
0
1
2
3
4
5
6
7
8
9
0 1000 2000 3000
CONSTRAINED MODULUS (bars)
Dep
th (
m)
0
1
2
3
4
5
6
7
8
9
0 2 4 6 8 10
MATERIAL INDEX ID
Dep
th (
m)
0
1
2
3
4
5
6
7
8
9
0 0.2 0.4 0.6
UNDRAINED SHEAR STRENGTH (bars)
Dep
th (
m)
0
1
2
3
4
5
6
7
8
9
0 10 20 30 40 50
HORIZONTAL STRESS INDEX KD
Dep
th (
m)
Figure 6-5. DMT sounding at Green Cove Springs.
219
0
2
4
6
8
10
12
14
16
-10 10 30 50 70 90 110
Volume (cm3)
Pres
sure
(bar
s) 1.0m
2.0m
3.0m
Figure 6-6. Corrected PMT curves at Green Cove Springs, depths 1.0 m, 2.0 m, 3.0 m.
0
2
4
6
8
10
12
14
16
-10 10 30 50 70 90 110
Volume (cm3)
Pres
sure
(bar
s) 5.0m
7.0m
10.0m
Figure 6-7. Corrected PMT curves at Green Cove Springs, depths 5.0 m, 7.0 m, 10.0 m.
220
Static Load Testing
Circular Concrete Footing
A circular 6.0 ft (1.8 m) diameter 2.0 ft (0.6 m) thick circular concrete footing
was made. In order to overcome the superficial crust at the site, the footing was
embedded 2.0 ft (0.6 m) in the ground. The soil-structure profile and plan views are
sketched in Figure 6-8, along with the position of the ground water table at depth 5.5 ft
(1.7 m),.
Figure 6-8. Soil-concrete footing and loaded area.
GWT5.5ft
2.0ft
0.0
6.0ft
GWT
Static Load
Loaded Area
52in. diametrerof loaded area
221
The two-foot thick concrete footing was designed to be a rigid footing for the load
testing; steel reinforcing rods were installed at both the top and the bottom. For the
concrete material, compressive strength of about f’c = 7000psi was targeted. During the
casting of the footing, four cylindrical concrete specimens of two different sizes, 6in.
diameter, 12in. high and 4.35in. diameter and 8in. high, were taken. After nearly 80 days
of age, the concrete specimens were tested for the unconfined compressive strength
according to the ASTM Standard C873-99: the standard method for Compressive
Strength of Concrete Cylinders Cast in Place in Cylindrical Molds. The unit weight of
the cylindrical specimens were also estimated using the ASTM standard: C138: Standard
Test Method for Density (Unit Weight). Table 6-1 presents the different magnitudes of
the concrete properties from using the Tinius Olsen device. The average compressive
strength for the three specimens turned out to be 6912psi, while the average unit weight
for the four specimens is 141.7pcf. During the crushing, the initial portion of the loading
was load-controlled where the elastic behavior of the concrete is assumed. At about 50%
of exepected yield force Fmax, the loading was changed to deformation-controlled. This
loading process is illustrated in Figure 6-9.
The steel reinforcing rods were determined according to the American Concrete
Institute (ACI) code for structural slabs of uniform thickness. The minimal amount
ρmin = 0.002 was followed for the top layout and for the bottom layout. Thus, for the
6.0 ft (1.8 m) diameter and 2.0 ft (0.6 m) thick footing, reinforcing rods made of 6 Rebar
#6 (Grade 60: Yield strength of fy = 60ksi) were used in two perpendicular directions, as
shown in Figure 6-9. Photographs of the footing before and after the casting of the
concrete are presented in Figure 6-11.
222
F (lb) Fmax
Deformation Controlled
Load Controlled LoadingAssumed elastic behavior
Def. (in)
Table 6-1. Physical and mechanical properties of the concrete specimens
Concrete Specimens crushing withTINIUS OLSEN, Model CMH 289 Controller (2000 Recorder)
Specimen No. 1 2 3 4 Average Diameter (in.) 6.15 6.10 4.36 4.35 5.24 Height (in.) 11.88 12.00 8.00 8.00 9.97 Dry weight (lb) 27.73 27.89 8.25 8.34 18.05 Submerged (lb) 15.40 15.64 4.62 4.68 10.08 Unit weight (pcf) 140.4 142.0 141.7 142.3 141.6 Fmax (lb) 220279 195634 xx 98474 171462 Stress (psi) 7415 6694 xx 6626 6912 Loading rate (lb/min) 60000 60000 26600 26600 43300 Def. rate (in./min) 0.038 0.03 0.02 0.02 0.03 Age (days) 80 80 80 80 80
XX: Undefined
Figure 6-9. Loading process using Tinius Olsen model CMH 289 controller.
Figure 6-10. Concrete footing with reinforcing steels and loaded area.
223
Figure 6-11. Footing before and after the casting of concrete.
Load Testing Equipments
The static loading on the footing was achieved by successively stacking Statnamic
dead weights. The magnitude of load and the corresponding settlement were successfully
monitored using three identical load cells and four LVDTs (evenly distributed on the
footing), respectively.
Figure 6-12. Load cells, LVDTs, bottom steel plate, and the footing.
224
These devices were installed on the footing as indicated in Figure 6-12, and
connected to a personal computer equipped with a special data acquisition system
MEGADAC to record the loads and displacements.
The MEGADAC data acquisition was set up to take the readings every 300
seconds. The three load cells were evenly placed on a steel plate (angle 120° apart from
each other about the center), Figure 6-12. The 52 in. diameter and 6 in. thick steel plate
was necessary in order to assure a leveled surface before the actual loading starts. The
load cells were then covered with a 5.5 ft diameter steel plate to accommodate the first
circular weight, Figure 6-13. The total weight from the bottom steel plate, the top steel
plate, and the seating weight is given below
Wbottom plate = 3,500 lbWtop plate = 5,440 lb ⇒ total seating weight: Wtotal = 16,140 lb (6.1)WSeating weight = 7,200 lb
Figure 6-13. Total seating weight on Green Cove Springs footing.
225
Photographs matching the descriptions of Figures 6-12 and 6-13 were taken and shown in
Figure 6-14.
Figure 6-14. Photographs for Load cells, LVDTs, steel plates, and seating weight.
Figure 6-14 also shows the reference beams used during the load testing. It is made of
fiberglass material, and was fabricated by the University of Florida, Geotechnical Group.
The main role of the reference beam is to support the four LVDTs. The two reference
beams are 24.0 ft (7.3 m) long and 3.0 ft (0.9) high, and were positionned symmetrically
(in two perpendicular directions) about the footing, about 7.0 ft (2.1 m) apart from each
other, Figure 6-15. Two secondary beams were attached to the principal beams in order
to support the two LVDTs located in the middle of the principal beams. Four posts,
which were 24.0 ft (7.3 m) apart longitudinally and 7.0 ft (2.1 m) apart laterally support
the hanging reference beams. Several small crossbeams were also needed (not included
in the figures) to increase the lateral stiffness of the two principal beams.
226
As mentioned above, the load cells and the LVDTs were hooked up to the
MEGADAC computer system. The data reading was started after the seating weight in
Figure 6-13 was installed.
Figure 6-15. Principal reference beam, poles and concrete footing.
The output screen instantly displayed the data from the load cells and the LVDTs.
Each time a new load (weight) was applied, its magnitude could be checked on the
MEGADAC computer output screen. Figure 6-16 is a drawing of the hook-up for the
data acquisition system, which is then shown in the photograph of Figure 6-17.
227
Figure 6-16. Data acquisition system and MEGADAC computer.
Figure 6-17. Data acquisition system at Green Cove Springs.
Static Loading Process
After all of the accessory equipment (reference beams, load cells, LVDTs,
MEGADAC acquisition computer, seating weights) were completely set up, the initial
readings in load and displacement were recorded; subsequently, the actual loading test
was started. Table 6-2 presents a summary of the schedule followed during the static
load testing. The dates, times, weights applied, and the cumulative pressures were given
for each day of load increment.
Load Cells 1, 2, 3 LVDTs 1, 2, 3, 4
LVDT 1 = ……...LVDT 2 = ………LVDT 3 = ………LVDT 4 = ………
Load Cell 1 = ………Load Cell 2 = ……..Load Cell 3 = ……..Output Screen
Increasing Load
FOOTING MEGADAC
228
Table 6-2. Loading schedule followed at Green Cove Springs
Partial Time Cumulative Time Weights Load Pressure*Date (sec) (sec) (days) (lb) (kips) (tsf)
6/27/01 0 0 0.00 6,600 5,500 1.21E+01 0.43
6/28/02 71100 7.11E+04 0.82 11,000 10,800 3.86E+01 0.90
6/29/01 81900 1.53E+05 1.77 9,670 9,559 9,075 6.69E+01 1.40
7/2/01 254100 4.07E+05 4.71 10,000 10,154 10,198 9.73E+01 1.93
7/3/01 85800 4.93E+05 5.70 8,877 10,132 10,022 1.26E+02 2.45
* Includes the weight of the lower steel plate and the footing
The illustration of Table 6-2 is provided in Figure 6-18 below and presented in
photographs in Figure 6-19. Finally, the plot of load versus elapsed time from the
MEGADAC computer for several hours after each load increment was put in place, is
presented in Figure 6-20.
Load on 6/27/01 Load on 6/28/01
Figure 6-18. Loading steps with weights at Green Cove Springs.
229
Load on 6/29/01 Load on 7/02/01
Load on 7/03/01
Figure 6-18. continued.
230
Figure 6-19. Installation of weights and final load applied.
0.00E+00
2.00E+01
4.00E+01
6.00E+01
8.00E+01
1.00E+02
1.20E+02
1.40E+02
0.00E+00 5.00E+04 1.00E+05 1.50E+05 2.00E+05 2.50E+05 3.00E+05
Cumulative Time (sec)
Tota
l Loa
d (k
ips)
Figure 6-20. Installation of weights and final loading state.
231
Survey Measurements
For verification and redundancy of the settlements recorded from the LVDTs, an
engineering survey of the load testing was undertaken simultaneously. In this study, the
engineering survey consists of mounting two scaled rulers on stationary objects that serve
as fixed references (back-sights) and two other scaled rulers (fore-sights) on objects that
are expected to settle as a result of the loading process. Locations of the different stations
during the survey operation are indicated in Figure 6-21. For the stationary back-sight
targets, a wall of the adjacent building and a stockpiled cylindrical weight were chosen;
while two diagonally opposite reference beam poles and the seating weight on the footing
were monitored for settlement during the loading process. The survey level was installed
at an optimal location in order to get the best reading accuracy from all of the targets.
The distances shown in Figure 6-21 are only indicative and not exact. The heights where
the five ruler targets were placed are in the neighborhood of 5.0 ft above the ground
surface, approximately the height of the level during the reading operations.
Modeling with CSANDSET and FEM code PLAXIS
Unlike the sheet pile unloading cases in the previous chapters, knowledge of the
stress history is very critical in loading of shallow footings. By difference of stress paths
in unloading and loading, the loading would cause the cap in the Hardening Soil model to
expand before it reaches the fixed failure surface. The stress history of subsurface soil in
Green Cove Springs was extensively studied.
Correlations for Soil Properties
Despite the stress path difference in unloading cases (sheet piles) and loading
cases (shallow footings), the correlations used for the theoretical predictions are the same
as those listed in Chapter 4.
232
Level
Stationary Weight
Stationary Wall
Pole 1
Pole 2Seating Weight
Ruler 1
Ruler 2
Ruler 3 Ruler 5
Ruler 4
30.0ft
30.0ft
30.0ft20.0ft
15.0ft
Adjacent Building
AFT
, Inc
., m
ain
offic
e w
all
Figure 6-21. Survey level and targets locations.
233
Only the correlations for the DMT data are discussed in this section. The main input
parameters (Chapter 3) are repeated below for convenience.
For sand layers (ID > 1.8), the fiction angle φ from the Dilatometer program is the
plane strain friction angle; it is evaluated from Eq. (6.2) (Marchetti et al., 2001)
( ) Hf FZSRODWTqDFRICBDFRICDMAREA
uRODDIAMTHRUST
+×+×
×−+
−××−=
24
019.142
tan
2
02ps
π
πφ
(6.2)
For clay layers (ID < 2), the undrained shear strength su in [psf].
( ) 25.10 5.022.0 Dvu Ks ×= σ (6.3)
where 0vσ is the corresponding effective overburden stress.
The elastic modulus E is calculated from the constrained modulus M, this latter is
obtained from the Dilatometer modulus ED.
( ) ( )( ) ( ) DEME ×−=×
−−×+
= 211
211 νν
νν (6.4)
where ν is the Poisson’s ratio.
For the finite element analysis, the estimation of the elastic modulus in case of an
overconsolidation sand is discussed below
Overconsolidation Ratio for Sands
The DMT was the insitu test that provides a good estimate of the initial stress
state parameters. Interpretation of the coefficient of lateral stress KD versus depth (Figure
6-5) is a practical tool to identify whether a layer is normally consolidated or
overconsolidated. The availability of CPT data from the same site helps estimate the
overconsolidation ratio, OCR, following the method proposed by Marchetti et al. (2001).
234
The method consists of using the ratio MDMT/qc of the sand layer as a reference, then for
values of MDMT/qc = 5-10, the sand is a normally consolidated sand; and for MDMT/qc =
12-24, it is an overconsolidated or cemented sand. Table 6-3 indicates that the sand layer
at the test site Green Cove Springs is an overconsolidated or cemented sand to a depth 9.2
ft (2.8 m).
The Overconsolidation Ratio, OCR, was computed with Dilatometer program,
(Schmertmann, 1983), however the values were considered to be high for a broad range
of soil types. Thus, the more universal correlation developed by Kulhawy and Mayne,
(1982) is used in this study:
φ
φsin
1
0
sin1
−
=K
OCR (6.5)
Table 6-3. Stress history of sand layers by combining DMT and CPT
Depth MDMT qc
(m) (bars) (bars)Ratio
MDMT/qc
0.2 1212 120.3 10.070.4 363 110.75 3.280.6 1403 120.4 11.650.8 1385 139.6 9.921.0 1083 94.4 11.471.2 1644 110.6 14.861.4 2068 113.25 18.261.6 2246 103.1 21.781.8 1058 90.3 11.722.0 1388 82.7 16.782.2 1229 51.2 24.002.4 784 33.2 23.612.6 430 24.4 17.622.8 406 33.6 12.083.0 354 64 5.533.2 280 51.4 5.453.4 249 53.2 4.68… … … …
235
Laboratory consolidation tests on “undisturbed” samples were also carried out in
order to confirm the idea that the sand layer at Green Cove Springs site was
overconsolidated. Consolidometer ring samples were extracted from depths 2.5 ft
(0.8 m) and 3.5 ft (1.1 m) below the ground surface, Figure 6-2. Samples in deeper layers
could not be taken as the excavated soils caved in quickly.
The estimation of the OCR values from the consolidation tests is presented in
Figures 6-22 and 6-23. Soil disturbance in sands during the hand sampling is considered
to be the main cause of the difference between the estimates from the consolidation
method and the Kulhawy and Mayne (1982)-based analysis. A comparison of the OCR
values from the consolidation tests, Kulhawy and Mayne (1982), and Schmertmann
(1983) at depths 2.5 and 3.5 ft is presented in Table 6-4.
The OCR values selected for the finite element analysis are those from Kulhawy
and Mayne (1982): not only that the equation was based on several tests that cover a good
variety of soil from clay to gravel (Coduto, 1994), but also, in our comparison, the values
are the closest to the average values at the depths 2.5 ft (0.8 m) and 3.5 ft (1.1 m). Also,
the correlation for the elastic modulus EOCR, already introduced in Chapter 3, was used
for the constitutive model Mohr-Coulomb (Bowles, 1996)
OCREE NCOCR ×= (6.6)
ENC is the elastic modulus assuming a normally consolidated sand. The correlation of Eq.
(6.6) is particularly selected among others because the modulus should be a function of
the OCR value (here nonlinear), rather than just a multiple of the ENC value from a given
range of multiplier as it is proposed by other authors.
236
Oedometer 1 - Depth 2.5ft
0.304
0.354
0.404
0.454
0.504
0.554
0.604
0.654
0.704
0.754
0.1 1 10 100 1000 10000 100000 1000000log P (tsf)
Void
ratio
e
0
1
2
3
4
5
6
Coe
ffici
ent C
v (c
m2 /m
in)
e0 = 0.726
0.42e0
σ'v0 =0.14tsf σ'c = 1.8tsf
Figure 6-22. OCR value at depth 2.5 ft from consolidation test.
237
Oedometer 2 - Depth 3.5ft
0.304
0.354
0.404
0.454
0.504
0.554
0.604
0.654
0.704
0.754
0.1 1 10 100 1000 10000 100000 1000000log P (tsf)
Void
ratio
e
0
1
2
3
4
5
6
Coe
ffici
ent C
v (c
m2 /m
in)
0.42e0
e0 = 0.722
σ'v0 =0.20tsf σ'c = 1.1tsf
Figure 6-23. OCR value at depth 3.5 ft from consolidation test.
238
Table 6-4. Comparison of overconsolidation ratio values
Depth(ft)
LaboratoryConsolidation
Kulhawy andMayne (1982)
DilatometerProgram*
2.5 12.8 28.1 66.2
3.5 5.5 15.6 30.3 * Schmertmann (1983), based on Kulhawy and Mayne (1982)
The estimation of the friction angle from the PMT used the curve fitting method
with PLAXIS explained in Chapter 4. The plots of the curve-fittings at each depth of test
are reported in Appendix D; the Hardening Soil model produced better results than the
Mohr-Coulomb model. A typical curve-fitting result at Green Cove Spring is plotted in
Figure 6-24.
0
2
4
6
8
10
12
14
16
-20 0 20 40 60 80 100 120
Volume (cm3)
Pres
sure
(bar
)
Insitu
Plaxis (HS)
Figure 6-24. Curve-fitting with PLAXIS (Hardening Soil model).
239
Modeling with CSANDSET
As the footing in Green Cove Springs is circular, the model used was a square
footing having the same contact area as that of the actual circular footing; thus, the edge
of the square footing becomes B = L = 5.32 ft (1.6 m). The net bearing pressure at the
base of the footing, from Table 6-2 was evaluated at 2.3 tsf. Also, the ground water table
is located at 5.5 ft (1.7 m) below the ground surface. For Oweis (1979) method and
Schmertmann (1978) method, four layers were specified before the rigid layer at depth
26.0 ft (7.9 m). Figure 6-25 shows the sketch the geometrical input for the conventional
methods.
Figure 6-25. Soil layers for Oweis (1979) and Schmertmann (1978).
The main input parameters for CSANDSET are listed in Table 6-5. The K0 value
is the average over the corresponding layer from the DMT results. The soil properties of
RIGID LAYER
Layer 3
Layer 2
Layer 1GWT
5.32ft
2.0ft
5.5ft
8.00ft
12.88ft
23.00ft
Layer 426.00ft
2.3tsf
240
the sublayers designated in Figure 6-25 and required for Oweis (1979) and Schmertmann
(1978) methods are given in Table 6-6.
Table 6-5. Main input parameters for CSANDSET
Soil Properties of Soil overB = 5.32 ft below Base Input Values
Blow count (Blows/ft) 13
CPT end bearing (tsf) 102.1
Total unit weight (pcf) 116
Saturated unit weight (pcf) 120
Coefficient K0 at rest 2.16 (from DMT)
Depth of rigid layer (ft) 26.00
Ground water table (ft) 5.5
Poisson’s ratio 0.3
Table 6-6. Soil properties of sublayers for Oweis (1979) and Schmertmann (1978)
Layer Bottom (ft) γm (pcf) γsat (pcf) SPT N qc (tsf) K0
1 8.00 116 120 13 102.1 2.162 12.88 110 115 6 56.4 0.633 23.00 100 110 2 30.1 0.664 26.00 100 110 3 7.0 1.04
In addition to the traditional methods in CSANDSET, the Dilatometer method of
settlement calculation and the Pressuremeter method of settlement calculation were also
used to predict the settlement at Green Cove Springs.
Modeling with PLAXIS
For the finite element modeling, the circular footing is adequately simulated using
the axisymetric model. The concrete was modeled as a linear elastic nonporous material.
The edge of the concrete footing was frictionless, that is the reduction factor on the side
241
Rf = 0.01 (minimal value accepted), as there is no contact between the soil and the side of
the footing. The properties of the concrete are listed in Table 6-7.
Table 6-7. Properties of the concrete footing
Properties Values
Unit weight (pcf) 150
Elastic modulus (psi) 4.3×10+6
Material type Nonporous
The domain is discretized into 15-noded triangular elements, each with 12 stress
points. The width of the model is 12.0 ft and was selected based on the location of the
poles in the field, points where no settlements are expected to occur; the height of the
model 26.0 ft (8.0 m) was estimated from the assumed depth of influence 4 times the
diameter of the footing (6.0 ft). The geometric configuration shown in Figure 6-26
represents only half of the finite element domain.
Figure 6-26. Finite element modeling with PLAXIS (Unit: m).
242
The input parameters for the finite element modeling are presented in Table 6-8.
Those properties are the “normally consolidated” properties of the soil layers, that is, the
elastic moduli are obtained from the correlations mentioned in Chapter 4. For the Mohr-
Coulomb modeling, the modulus value is obtained from Eq. (6.6) if the layer in question
is overconsolidated, except in DMT. As for the Hardening Soil model, the OCR value
obtained from Eq. (6.5) is input in the program through the Initial Conditions (input sub-
stage) of each overconsolidated soil cluster (alternatively, the preconsolidation stress can
be evaluated by knowing the OCR and input in the Initial Conditions stage), and the ENC
values in Table 6-8 are kept as they are in the main input.
Soil Properties from Consolidation Test
Consolidation theory-based and finite element-based computations of the footing
settlement using the soil properties from the consolidation test were also performed.
Because of lack of data, the properties corresponding to the depth 3.5 ft was extended to
12.0 ft below the base of the footing (assumed to be the depth of influence). The soil
properties from the consolidation tests are listed in Table 6-9. The “Soft Soil model”,
commonly known as Cam-Clay model, was the consititutive model selected for PLAXIS,
and the soil layers were assumed to be in drained conditions.
Estimate of Bearing Capacity
As a part of the geotechnical investigation for this study, the bearing capacity of
the soil for footing was also estimated using the following three methods: Vesic (1975),
Meyerhof (1963), and Terzaghi (1943). The computations were performed in Excel
spreadsheet and are presented in Table 6-10.
243
Table 6-8. Soil properties from different insitu tests for PLAXIS (OCR = 1)
SPT CPT and CPT (N) DMT PMTBottom N γm φ Su E qc φ φ (N) Su E E (N) Μ φ Su E pl EM φ Su E Bottom
(ft) (bl/ft) (pcf) (o) (psf) (psi) (psi) (o) (o) (psf) (psi) (psi) (psi) (o) (psf) (psi) (psi) (psi) (o) (psf) (psi) (ft)
GS=0 214.6 2414.3 37.1 - 3567.9 GS=04.92 14 120.0 31.4 - 2102.5 1549.7 44.6 35.8 - 4649.0 2863.8 4.92
6.23 21933.1 46.3 - 16293.1 6.23
6.56 6.56
7.22 208.8 1905.3 33.4 - 2815.8 7.22
7.51 10 110.0 30.1 - 1812.5 7.51
8.20 594.5 36.6 32.4 - 2157.6 2138.8 8.20
9.51 6 100.0 28.6 - 1522.5 5895.3 41.0 - 4379.3 9.51
9.84 9.84
11.81 116.0 849.1 26 - 1254.8 11.81
13.06 10135.5 43.2 - 7529.3 13.06
13.12 1115.3 39.1 33.7 - 3346.0 2407.0 13.12
13.45 13.45
13.94 3326.7 37.1 - 2471.2 118.9 877.3 25 - 1309.5 13.94
17.72 17.72
19.69 2 90.0 27.1 - 1232.5 429.2 35.2 30.4 - 1133.5 1698.5 19.69
20.01 720.7 - 751.7 535.3 20.01
25.98 104.7 - 803.8 575.5 1116.7 55.1 152.3 - 751.7
* 227.225.98
27.89 27.89
32.81 32.81
36.09 65.3 207.5 - 867.8 204.5 36.09
EOB EOBThe correlations are those listed in Chapter 4 and present chapter
244
Table 6-9. Soil properties from consolidation test
Properties Layer 0.0 ft to 1.0 ft Layer 1.0 ft to 12 ft
Thickness (ft) 1.0 11.0
Void ratio, e0 0.726 0.722
Compression index, Cc 0.0622 0.0611
Swelling index, Cs 0.0149 0.0147
Preconsolidation stress (tsf) 1.8 1.1
Sublayers thickness, ∆H (ft) 0.2 (5 layers) 0.5 (22 layers)
Permeability (ft/day) 11.48×10-5 13.58×10-5
Table 6-10. Bearing capacity estimation for concrete footing
*Most conservative value of the friction angles (SPT)
γm (pcf) =117.7B (ft) =6.0
Df (ft) =2.0GWT (ft) =5.5γavg (pcf) =92.6
VESIC (1975) MEYERHOF (1963) TERZAGHI (1943)φ* (o) =31.4 φ (o) =31.4 φ (o) =31.4
Bearing Capacity FactorsNφ =3.18 Nφ =3.18Nq =21.61 Nq =21.61 Nq =26.52Nγ =5.10 Nγ =19.87 Nγ =25.35
Correction Factorsλqs =1.61 λqs =1.32λqd =1.09 λqd =1.06λqi =1.00 λqi =1.00λγs =0.60 λγs =1.32λγd =1.00 λγd =1.06λγi =1.00 λγi =1.00
Ultimate Bearing Capacitiesqult (tsf) =4.90 qult (tsf) =7.41 qult (tsf) =5.12
245
Settlement Results
Field Measured Settlements
The output data from the four LVDTs and the load cells are plotted versus time.
The reading was taken every 300 seconds (5 minutes) during the 6-day load testing
(Table 6-2). Only the settlements and loads recorded few hours after application of each
incremental load are presented in the output graph, Figure 6-27. As the readings from the
four LVDTs are not identical, but reasonably within the same range, the average value
was used for the analysis.
Figure 6-27. LVDTs settlements and load versus time.
The settlement observed from the survey operation is consistent with the average
values from the LVDTs. Table 6-11 shows the results from LVDTs and Survey for
-4.00E-02
-2.00E-02
0.00E+00
2.00E-02
4.00E-02
6.00E-02
8.00E-02
1.00E-01
1.20E-01
0.00E+00 5.00E+04 1.00E+05 1.50E+05 2.00E+05 2.50E+05 3.00E+05
Elapsed Time (sec)
Settl
emen
t (in
ches
)
0.00E+00
2.00E+01
4.00E+01
6.00E+01
8.00E+01
1.00E+02
1.20E+02
1.40E+02
Foot
ing
Load
(kip
s)
LVDT 1LVDT 2LVDT 3LVDT 4Total Load
246
comparison. It can be stated that the precision on the survey was satisfactory and the
LVDTs measurements did not show any signs of any technical problems.
Table 6-11. Settlements measured with LVDTs and survey versus loads
Loading Sequence (tsf)Settlement (in.)0 0.276 0.746 1.247 1.783 2.318
LVDT Measured (average) 0 0.00 0.02 0.04 0.05 0.10 SURVEY Measured 0 — — 0.06 0.12 0.20
Results from CSANDSET and Other Traditional Methods
The results from the traditional methods include those from CSANDSET, the
insitu test based methods: DMT and PMT methods, and the consolidation theory; they are
given in Tables 6-12 and 6-13 from the SPT N and the CPT (N) data, respectively.
Table 6-12. Settlements from SPT N and other traditional methods
Loading Sequence (tsf)Settlements (in.)0.276 0.746 1.247 1.783 2.318
LVDT Measured 0.00 0.02 0.04 0.05 0.10 Conventional Oedometer Test 0.29 0.49 0.65 0.85 1.10 PMT 0.04 0.13 0.24 0.35 0.46 DMT 0.04 0.11 0.19 0.26 0.34 Terzaghi (1948)a) 0.16 0.44 0.74 1.06 1.37 Teng 0.04 0.12 0.2 0.28 0.37 Elastic Theory b): Rigid 0.09 0.24 0.4 0.57 0.74 Center 0.09 0.26 0.43 0.61 0.80 D'Appolonia (1970) a) 0.03 0.09 0.16 0.22 0.29 Peck & Bazaraa (1969) a) 0.05 0.15 0.25 0.37 0.49 Schmertmann (1970)c) 0.05 0.18 0.32 0.47 0.62 Schmertmann (1978) c) 0.04 0.16 0.31 0.47 0.65 Schultz & Sherif (1973) a) 0.07 0.14 0.23 0.32 0.41 Meyerhof (1965) a) 0.08 0.21 0.35 0.50 0.65 Peck, Hanson, Thornburn a) 0.15 0.42 0.7 1.00 1.30 Bowles b) 0.13 0.36 0.6 0.86 1.12 NAVFAC DM 7.1 (1982) a), c) 0.10 0.26 0.44 0.63 0.82 Oweis (1979) a), b): Rigid 0.02 0.12 0.34 0.69 1.15 Center 0.03 0.23 0.64 1.28 2.08
a) SPT based Methodb) Elastic Theory based Methodc) CPT based Method
247
The blow count N from SPT and CPT produced rather significantly different
settlement predictions. Generally, using the CPT (N) appears to be more accurate than
the SPT N.
Table 6-13. Settlements from CPT (N) and other traditional methods
Loading Sequence (tsf)Settlements (in.)
0.276 0.746 1.247 1.783 2.318 LVDT Measured 0.00 0.02 0.04 0.05 0.10 Conventional Oedometer Test 0.29 0.49 0.65 0.85 1.10 PMT 0.04 0.13 0.24 0.35 0.46 DMT 0.04 0.11 0.19 0.26 0.34 Terzaghi (1948)a) 0.09 0.24 0.40 0.57 0.74 Teng 0.02 0.06 0.10 0.14 0.18 Elastic Theory b): Rigid 0.06 0.17 0.28 0.40 0.52 Center 0.07 0.18 0.30 0.43 0.56 D'Appolonia (1970) a) 0.03 0.07 0.12 0.17 0.22 Peck & Bazaraa (1969) a) 0.02 0.07 0.13 0.18 0.24 Schmertmann (1970)c) 0.04 0.18 0.31 0.46 0.59 Schmertmann (1978) c) 0.04 0.20 0.38 0.59 0.80 Schultz & Sherif (1973) a) 0.04 0.08 0.13 0.19 0.24 Meyerhof (1965) a) 0.04 0.11 0.19 0.27 0.35 Peck, Hanson, Thornburn a) 0.08 0.22 0.36 0.52 0.67 Bowles b) 0.07 0.19 0.33 0.47 0.60 NAVFAC DM 7.1 (1982) a), c) 0.06 0.16 0.26 0.38 0.48 Oweis (1979) a), b): Rigid 0.01 0.04 0.10 0.21 0.34 Center 0.01 0.07 0.20 0.39 0.62
a) SPT based Method b) Elastic Theory based Method c) CPT based Method
Results from PLAXIS
The settlements from the finite element predictions with PLAXIS are pesented in
Table 6-14. The results showed more consistency compared to those from the traditional
methods. Generally, the settlements from using the Mohr-Coulomb model were more
accurate than those from the Hardening Soil model. The settlements estimated from
using the Soft Soil model were the largest.
248
Table 6-14. Settlements from finite element analysis with PLAXIS
Loading Sequence (tsf)Settlements (in)
0.276 0.746 1.247 1.783 2.318 LVDT Measured 0.00 0.02 0.04 0.05 0.10 Soft Soil Model with PLAXIS 0.27 0.49 0.84 1.36 2.05
SPT 0.04 0.13 0.28 0.50 0.73 CPT (N) 0.04 0.11 0.21 0.35 0.52 CPT 0.04 0.12 0.21 0.31 0.41 DMT 0.04 0.12 0.22 0.32 0.43
MOHR-COULOMB
PMT 0.08 0.24 0.45 0.71 1.04 SPT 0.08 0.22 0.36 0.53 0.81 CPT (N) 0.06 0.17 0.28 0.42 0.61 CPT 0.07 0.19 0.34 0.52 0.72 DMT 0.06 0.17 0.29 0.42 0.57
HARDENINGSOIL
PMT 0.11 0.29 0.51 0.80 1.24
Results by Insitu Tests
The load-settlement curves from the LVDTs and all of the settlement predictions
are plotted by name of insitu test for better comparison; for unit conversion: 1 tsf =
95.8kPa. The plots are given in Figure 6-28 through Figure 6-32.
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 50 100 150 200 250
Applied Pressure (kPa)
Sett
lem
ent (
in)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 6-28. Predicted settlements using SPT data.
249
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 50 100 150 200 250
Applied Pressure (kPa)
Sett
lem
ent (
in)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 6-29. Predicted settlements using CPT (N) data.
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 50 100 150 200 250
Applied Pressure (kPa)
Sett
lem
ent (
in)
Measured
Schmertmann (1978)
Plaxis MC
Plaxis HS
Figure 6-30. Predicted settlements using CPT data.
250
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 50 100 150 200 250
Applied Pressure (kPa)
Sett
lem
ent (
in)
Measured
DMT Method
Plaxis MC
Plaxis HS
Figure 6-31. Predicted settlements using DMT data.
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 50 100 150 200 250
Applied Pressure (kPa)
Sett
lem
ent (
in)
Measured
PMT Method
Plaxis MC
Plaxis HS
Figure 6-32. Predicted settlements using PMT data.
251
Discussion of Results
At first glance, all of the calculation methods, traditional and finite element,
overpredict the settlement of the footing throughout the static load test. As indicated in
Tables 6-12 and 6-13, the traditional methods with CSANDSET are primarily based on
the SPT N blow counts and the CPT tip resistance qc. For this analysis, given that the
CPT (N) based method results are, by far, more accurate than those from the SPT N,
mostly the CPT (N) data and results are brought into the discussion.
The most accurate final settlement was estimated with D’Appolonia (1970)
method, 0.22 in.; this method evaluates the modulus of compressibility of the sand from
the SPT N blow counts, with some correction factors. The next most accurate calculation
methods are Schultz & Sherif (1973) method and Peck & Bazaraa (1969) method, both
with the final settlement of 0.24 in.. The actual settlement is so small that even a
relatively small overprediction appears to be too excessive in the problem. In fact the
three most accurate calculation methods above are at least twice as much as the final
measured settlement. Nevertheless, the best prediction settlement is still within a margin
0.2 in. from the actual settlement.
On the other hand, Schmertmann (1978) method is the least accurate with 0.84 in.
of total settlement. Schmertmann (1978) method, based on the CPT, is considered to be
among the most conservative methods in predicting settlement of shallow footings in
sands. Following Schmertmann (1978) is Terzaghi (1948) method with a total settlement
of 0.74 in.. Terzaghi (1948) method is also based on the SPT N. The conventional
methods based on the DMT and PMT also give promising settlement predictions. The
PMT method that is mainly based on the pressuremeter soil strength and stiffness
parameters (pl and EM) is in fairly good accuracy with 0.46 in.. The DMT method, which
252
uses the constrained modulus MDMT, results in a closer estimation with 0.34 in.. The
reason the difference between the actual settlement and the computed settlements is not
completely clear but could be attributed to the estimation of the correct overconsolidation
ratio OCR. The DMT results showed that the overlying sand is overconsolidated but the
Kulhawy and Mayne (1982) correlations seem to underestimate the OCR value in this
problem. The OCR values from the Dilatometer program, Schmertmann (1983) was not
used for safety reason and less number of datapoints.
In the finite element analysis with PLAXIS, the input data that resulted in best
accuracy of the calculation are those from the CPT and the DMT. Generally, the Mohr-
Coulomb model predicts better than the Hardening Soil model does. In the case of DMT
data, the elastic modulus of the overconsolidated layers is directly obtained from the
values of the constrained modulus MDMT, whereas in the SPT, CPT and PMT, the
procedure described earlier in this section (product of the square root of the OCR value
and the normally consolidated modulus). The settlement predicted from the CPT was the
smallest. This is estimated as result of the combination of slightly higher friction angles
and application of Eq. (6.6) for the modulus, Table 6-8. The other insitu tests settlement
predictions: SPT, CPT (N), and the PMT, did not give good settlement predictions. The
consolidation theory-based input parameters (laboratory test) and the PMT input
parameters provided the least accurate results with PLAXIS, followed by the SPT data,
2.05 in., 1.04 in. and 0.73 in., respectively. The SPT N values are too small considering
that the first sand layer is overconsolidated. The equivalent CPT (N) values confirmed
that the more consistent N values are larger than the SPT N values. This unreliability of
the SPT data is not an uncommon problem. It has been explained in Chapter 3 that the
253
SPT does not provide good reproducibility of data. Based on observation in the field
during the SPT, the SPT data could be affected by the operator and equipment sensitivity
(energy) factors. The input parameters from the CPT (N) have a reasonably accurate
settlement result among the finite element analyses, 0.52 in.. This prediction however is
about 5 times the actual settlement. The Hardening Soil model with PLAXIS generally
overpredicted the settlements; the smallest prediction was 0.57 in. from the DMT and the
largest prediction was 1.24 in. from the PMT. Nevertheless, the Hardening Soil model
seems to provide less discrepancy of the results with the different insitu testing input
parameters (omitting the PMT result) than the Mohr-Coulomb model.
Conclusions
Results from the Green Cove Springs footing were very useful for primarily
understanding which insitu test best solves the problem, when to use traditional methods
or finite element methods, and which model to use in the finite element analysis with
PLAXIS. The existence of the overconsolidated sand layer within the zone of influence of
the footing concludes that a careful evaluation of the stress history of a site was critical in
the settlement predictions. The important parameters for the stress history were the
coefficient of lateral pressure K0 and the overconsolidation ratio OCR. It is rather
difficult to estimate from the laboratory consolidation test the OCR value (alternatively,
the past maximum stress) of sand due to soil disturbance when sampling. The usefulness
of the DMT in this regard has been very evident. The DMT was the insitu testing that
provides for the initial stress state of a soil profile without resorting to correlations. It has
been realized that the availability of CPT data from the same site helps to better
understand the stress history of a soil profile. The following conclusions are obtained:
254
It is fundamental to know the best interpretation of the stress history of a soil
profile, in this case sand, for the settlement prediction either using conventional methods
or finite element methods;
The conventional methods, especially D’Appolonia (1970), Schultz & Sherif
(1973), Peck and Bazaraa (1969), and DMT method provide more accurate predictions in
overconsolidated sands;
The finite element method using DMT data with Mohr-Coulomb model is more
reliable; no evaluation of the OCR value is required as the constrained elastic modulus
MDMT is considered to incorporate the stress history thorugh the coefficient of lateral
stress KD;
The CPT is also a good method for the case of settlement predictions of shallow
footings in sands, the higher friction angle values make up for the more accurate results;
The N blow count correlated from the CPT data is more consistent with the other
insitu test data than the SPT N blow count for both traditional methods and finite element
method with PLAXIS.
The PMT and the SPT do not provide good settlement predictions in case where
the sand layer within the zone of influence is overconsolidated.
The Hardening Soil model in finite element model overpredicts the settlement in
overconsolidated sands using any type insitu input parameters.
The following recommendations can also be stated:
The DMT and CPT insitu tests are mandatory in evaluating settlement of a
shallow footing on sands. Not only do they give better predictions but also help define the
stress history of the soil profile.
255
The SPT and PMT are not reliable in predictions of settlements using either
traditional methods or finite element methods.
It is less tedious and more accurate to predict the settlement using the traditional
methods than using the finite element method.
If the finite element method is the choice in settlement prediction, the Mohr-
Coulomb model provides better results than the Hardening Soil model when used with
the OCR value taken into account.
256
CHAPTER 7STUDY OF OTHER SHALLOW FOOTING CASES
General
This chapter deals with the analyses of other shallow footings on granular soils, to
confirm the results obtained from the Green Cove Springs footing in Chapter 6. A total
of three shallow footings were included: two concrete square footings from the FHWA1
project entitled Five Spread Footings on Sand in Texas A&M University, Texas, 1994;
and a circular chimney foundation in Intermountain Generating Station in west central of
Utah, 1986.
Two Square Concrete Footings in Texas A&M University
Introduction
The project on five spread footings is a full-scale static load tests sponsored by the
FHWA. The main objective was to re-evaluate the design and performance of spread
footings as alternative to the use of pile foundations for highway bridges. The square
spread footings were of different sizes ranging from 1.0 m×1.0 m to 3.0 m×3.0 m. For
this research, one footing size 2.5 m×2.5 m and one footing size 3.0 m×3.0 m were
selected. All of the footings are embedded 0.76 m in the ground and are 1.2 m thick.
Before the actual load tests, several insitu tests and laboratory tests were conducted
between November 1992 and May 1993 for extensive site exploration.
1 Federal Highway Administration
257
Site Description
A major advantage of analysing this project was the availability of results from
extensive soil insitu tests conformed to the ASTM standards. Among many other insitu
tests, 6 SPT with evaluation of energy efficiency, 5 CPT, 4 DMT, and 4 pre-bored PMT
were performed. The locations of the soundings relative to the footings under study are
displayed in the ASCE Special Publication No. 41 by Briaud and Gibbens (1994). Based
on the proximity ciriterion, only some of the insitu test data were used; for the DMT data,
only the sounding DMT3 was used because of availability of the Thrust data (the DMT4
is on far south of the plan view and is not included).
The borings SPT2 and SPT3 were selected for the analysis of Footing
3.0 m×3.0 m whereas SPT2 and SPT5 for Footing 2.5 m×2.5 m. Clearly, CPT5 would be
the best CPT data for Footing 3.0 m×3.0 m while CPT2 for Footing 2.5 m×2.5 m. PMT1
and PMT2 were selected for both footings, as they are the most complete data among the
four PMT soundings. More details on the insitu tests and laboratory tests are provided in
the Geotechnical Special Publication edited by Briaud (1994). The soil profile down to
the depth 11.0 m is predominantly sand formed in a coastal plain environment (middle
Eocene sand); after 11.0 m, a hard, plastic and dark clay layer exists. The sand layer is
essentially medium dense fine sand. The typical plots of the insitu basic parameters are
presented in Figures 7-33 and 7-34 for the SPT and DMT, respectively, and in Briaud and
Gibbens (1994) for the CPT.
The data reduction for CPT2 and CPT5 are reported in Appendix E. The PMT
curves from PMT1 and PMT2 are also given in Appendix E; for indications, only the
strength and the stiffness parameters from the sounding PMT1 are presented in Table 7-1.
258
SPT Blowcount Profile
0
1
2
3
4
5
6
7
8
9
10
11
12
13
14
15
16
0 10 20 30 40 50 60 70 80 90 100
SPT N
Dep
th (m
)
SPT1 SPT2 SPT3 SPT4 SPT5
Figure 7-33. SPT N versus depth
The soil profile in Figure 7-34 shows that the sand layer is somewhat uniformly
overconsolidated; further analysis will evaluate the overconsolidation ratio OCR values.
The overconsolidation may be due to dessication of the sand fines and removal of the
superficial layer of variable thickness between 0.5 and 1.5 m before the beginning of the
works (Briaud, 1994).
259
0
1
2
3
4
5
6
7
8
9
0 1 2 3 4
Material Index I D
Dep
th (m
)
0
1
2
3
4
5
6
7
8
9
0 1000 2000
Constrained Modulus M (bar)
Dep
th (m
)
0
1
2
3
4
5
6
7
8
9
0 1 2 3 4 5
Undrained Shear strength s u (bar)
Dep
th (m
)
0
1
2
3
4
5
6
7
8
9
0 5 10 15 20
Horizontal Stress Index K D
Dep
th (m
)
Figure 7-34. DMT3 sounding versus depth in Texas A&M University.
260
Table 7-1. Soil properties from PMT1 and PMT2
E0 (kPa) ER (kPa) pl (kPa)Depth (m)
PMT1 PMT2 PMT1 PMT2 PMT1 PMT2
0.6 6421 7621 21904 34336 400 580
1.2 6270 10558 24335 47885 460 900
2.1 7443 8075 33903 57779 800 840
3.4 7746 9490 51428 55513 740 920
5.2 12595 13365 78108 84967 1100 1250
7.6 4619 6387 61099 41970 900 1000
11.0 173878 132929 369548 253725 4200 4400
Test Setup
The footings were loaded with a jack reacting against a girder anchored into the
ground with diwidag bars, the latter being embedded in belled reaction shafts of 0.91 m
in diameter (anchored reaction system). Four LVDTs were placed at the corners of each
footing for the measurement of the settlements. The applied load was measured with a
calibrated load cell with a maximum capacity of 12,000 kN. Other measurement devices
used wire inclinometers, to monitor the movement of surrounding soils; and telltale to
measure the settlements in the ground. The reference beams were made of steel; their
movements were also monitored and the footing final settlements were adjusted
accordingly. The load test setup can be found in Briaud and Gibbens (1994).
Each load increment was applied for 30 minutes, but the footing settlement
readings were taken at 1, 3, 5, 7, 10, 15, 20, 25, and 30 minutes. The loading mode was
strain controlled, and each footing was load tested to the maximum settlement of 150
mm.
261
Modeling and Predictions
The properties of the footing reinforced concrete were taken to be the same as
those listed in the Green Cove Springs footing, Table 6-7. Regarding to the footings
sizes and thickness, assumption of rigid footings would be more appropriate for the
analyses. The most important properties of the sand, which is the overconsolidation ratio
OCR values is discussed next.
Overconsolidation ratio of sand layer
Because of the completeness of the insitu data from the Texas A&M load test
project, a comparative study on the estimation of the overconsolidation ratio OCR of the
sand was attempted. The evaluation of the OCR values with the correlation, Eq. (3.32)
from Kulhawy and Mayne (1982) was used in chapter 6. Estimation of the OCR values
from DMT combined with the CPT data, and from the CPT data alone would also be of
great interest. The OCR values are all estimated from the correlation developed by
Kulhawy and Mayne (1982), Eq. (3.32). It is the coefficient of lateral stress at rest K0
that is determined from different methods. In addition to the evaluation of the OCR
values from the DMT K0 using Eq. (3.32) (Chapter 6), the following four methods were
examined:
• Estimation of K0 using the DMT/CPT combination method developed by Baldi et al.(1986) to calculate the OCR value with Eq. (3.32).
• Estimation of the relative density Dr of the sand using Eq. (3.29) developed byJamiolkowski et al. (1985), then calculating the magnitude of the effective horizontalstress 0hσ from Eq. (3.31), then the K0 values to obtain the OCR values fromEq. (3.32).
• Estimation of the relative density Dr from the simplified form of Eq. (3.30) foroverconsolidated sands, (Kulhawy and Mayne, 1990), using an evaluation of 0hσfrom Eq. (3.31), and also using the K0 values to obtain the OCR values from Eq.(3.32), this is the method proposed (for sands) and used in this research.
262
• Iterative method proposed by Hortvah in 1994 (Briaud, 1994) for the same settlementprediction symposium.
Firstly, the repeatability of the CPT data was checked by plotting the OCR values
versus depth from the Baldi et al. (1986) method with DMT3/CPT5 and DMT3/CPT2.
The plots, shown in Figure 7-35 exhibit a very good consistency between CPT5 and
CPT2, and good reproducibility of OCR values from the DMT/CPT combination method.
Baldi et al . (1986)
0
1
2
3
4
5
6
7
8
9
0 2 4 6 8 10
DMT/CPT OCR D
epth
(m)
CPT5 CPT2
Figure 7-35. Estimation of OCR values from DMT/CPT data.
Secondly, the OCR values from the different methods are compared in Figure7-
36. Good agreement of the data was obtained except those from the Jamiolkowski (1985)
method where the OCR values are rather too small; as expected, the DMT K0 method
resulted in largest values; the method proposed in this research and the iterative method
proposed by Hortvah (1994) produced similar OCR values and variation along the depth.
263
In terms of arithmetic mean (rather than harmonic mean because of the more uniform
character of the OCR values versus depth) over the thickness 9.0 m, the minimum OCR
value was from the Jamiolkowski (1985) method: OCR = 0.83 (CPT5), followed by the
Baldi et al. (1986) and Hortvah (1994) methods: 3.33 and 3.23 from CPT5 data, and 3.49
and 3.55 from CPT2 data, respectively. The largest arithmetic mean was from the DMT
K0 values with Kulhawy and Mayne (1982): 5.81.
DMT3 and CPT5 data
0
1
2
3
4
5
6
7
8
9
0 2 4 6 8 10DMT3 - CPT5 OCR
Dep
th (m
)
Baldi et al. (1986) Method proposedHortvah (1994) K & M (1982)Jamiolkow ski (1985)
DMT3 and CPT2 data
0
1
2
3
4
5
6
7
8
9
0 2 4 6 8 10DMT3 - CPT2 OCR
Dep
th (m
)
Baldi et al. (1986) Method proposed
Hortvah (1994) K & M (1982)
Figure 7-36. Overconsolidation ratios of sand layer.
The OCR values from the method proposed in this research represent intermediate values
for the above range, with the arithmetic means: 3.53 from CPT5 and 3.74 from CPT2.
The method is considered to be slightly more conservative (but not too conservative) than
the method used in Chapter 6.
264
Soil properties
Several soil profiles were derived from the various insitu tests and borings. As a
summary of the insitu tests selected for each footing, Table 7-2 was established for quick
reference on the selected insitu test borings and the method of calculation for the
settlement predictions.
Table 7-2. Insitu tests and calculation methods for each footing
FOOTING
Dimension 3.0m×3.0m 2.5m×2.5m
Insitu test borings CPT5, SPT3, SPT2, PMT1 CPT2, SPT3, SPT2, PMT1
Insitu traditional methods DMT3, DMT1, PMT1, PMT2, DMT3, DMT1, PMT1, PMT2
CSANDSET methods SPT3, SPT2, CPT5 SPT5, SPT2, CPT2
Finite element method PLAXISSPT2, SPT3, CPT5 (N), CPT5,DMT3, PMT1, PMT2
SPT2, SPT5, CPT2 (N), CPT2,DMT3, PMT1, PMT2
The hard clay layer at depth 11.0 m is taken as the Rigid Layer in the traditional methods
with CSANDSET. The soil properties for the various calculations for each footing are
given in the next subsections.
Footing 3.0 m×3.0 m. Table 7-3 through Table 7-9 contain the soil properties
used for settlemment predictions of Footing 3.0m×3.0m. The correlations are those listed
in Chapters 6 and 7.
265
Table 7-3. Soil properties from SPT2 for both footings
Bottom(m)
γ(kN/m3)
SPT N(bl/ft)
φ(o)
su(kPa)
E(kPa) OCR Eocr
(kPa)GS = 0.0
0.6017.6 12.0 30.5 - 8500.0 10.00* 26879.4
1.2019.3 23.0 33.9 - 14000.0 6.44 35518.1
9.7518.5 18.0 32.4 - 11500.0 3.29 20871.8
15.0019.5 49.8 - 483.2 241578.9 1.00 241578.9
* extapolated from logarithmic variation of OCR at the top
Table 7-4. Soil properties from SPT3 for footing 3.0m×3.0m
Bottom γ SPT N φ su E Eocr
(m) (kN/m3) (bl/ft) (o) (kPa) (kPa) OCR (kPa)GS = 0.0
1.2018.2 15.0 31.6 - 10250.0 6.44 26004.3
1.8019.6 25.0 34.4 - 15000.0 2.77 24951.3
4.2018.5 18.0 32.4 - 11500.0 3.03 20027.8
8.2519.2 22.0 33.7 - 13666.7 3.61 25960.2
9.7517.3 10.0 29.9 - 7500.0 3.73 14490.3
15.0020.3 56.0 - 526.1 263065.2 1.00 263065.2
Table 7-5. Soil properties from CPT5 (N) for footing 3.0m×3.0m
Bottom CPT (N) φ su E Eocr
(m) (bl/ft) (o) (kPa) (kPa) OCR (kPa)GS = 0.76
9.5022.0 33.8 - 13500.0 4.05 27162.9
10.2550.0 40.2 - 27500.0 2.18 40620.6
15.2533.0 - 300.0 45000.0 1.00 45000.0
266
Table 7-6. Soil properties from CPT5 for footing 3.0m×3.0m
Bottom qc φ su E Eocr
(m) (bar) (o) (kPa) (kPa)OCR
(kPa)GS = 0.76
6.2593.70 40.7 - 28111.4 4.51 59731.1
9.5060.49 36.8 - 18147.7 2.52 28785.7
10.25261.50 43.6 - 78450.0 2.18 115879.5
15.0087.98 - 545.0 48390.8 1.00 48390.8
Table 7-7. Soil properties from DMT3 for both footings
Bottom M φ E(m) (bar) (o) (kPa)
GS = 0.0
2.40625.5 39.8 42216.6
3.80876.0 37.9 59123.5
5.401177.3 36.8 79455.6
6.80846.3 36.7 57118.0
8.201460.9 35.2 98597.0
Table 7-8. Soil properties from PMT1 for both footings
Bottom γ EM pl Rheo. Fac. Es = EM/α φ* sU Eocr
(m) (kN/m3) (kPa) (kPa)EM/pl
α (kPa) (o) (kPa)OCR
(kPa)GS = 0.0
0.9018.1 6421 400 16.05 0.50 12842 43.0 - 10.00 40610.0
1.6519.1 6270 460 13.63 0.50 12540 32.0 - 4.65 27047.6
2.7518.5 7443 800 9.30 0.33 22329 32.0 - 3.04 38919.2
4.3018.7 7746 740 10.47 0.33 23238 31.0 - 2.97 40056.4
6.4019.5 12595 1100 11.45 0.33 37785 31.0 - 3.24 67966.1
9.3018.0 4619 900 5.13 0.25 18476 28.0 - 3.93 36647.2
11.0021.0 173878 4200 41.40 1.00 173878 - 492.3 1.00 173878.0
* from curve fitting, see Appendix E
267
Table 7-9. Soil properties from PMT2 for both footings
Bottom γ EM plRheo.Fac. Es = EM/α φ* sU Eocr
(m) (kN/m3) (kPa) (kPa)EM/pl
α (kPa) (o) (kPa)OCR
(kPa)GS = 0.0
0.9018.1 7621 580 13.14 0.50 15242 43.0 - 10.00 48199.4
1.6519.1 10558 900 11.73 0.33 31674 40.0 - 4.65 68317.9
2.7518.5 8075 840 9.61 0.33 24225 32.0 - 3.04 42223.9
4.3018.7 9490 920 10.32 0.33 28470 31.0 - 2.97 49075.0
6.4019.5 13365 1250 10.69 0.33 40095 31.0 - 3.24 72121.2
9.3018.0 6387 1000 6.39 0.25 25548 28.0 - 3.93 50674.6
11.0021.0 132929 4400 30.21 1.00 132929 - 523.1 1.00 132929.0
* from curve fitting, see Appendix E
Footing 2.5 m×2.5 m. Some of the soil properties for Footing 2.5 m×2.5 m were
provided in the previous subsection (SPT5, DMT3, PMT1 and PMT2), the remaining soil
profiles are given in Tables 7-10 through 7-12. The correlations listed in Chapters 6 and
7 were used.
Table 7-10. Soil properties from SPT5 for footing 2.5m×2.5m
Bottom γ SPT N φ su E Eocr
(m) (kN/m3) (bl/ft) (o) (kPa) (kPa)OCR
(kPa)GS = 0.0
1.2017.8 13 30.9 - 9000.0 6.44 22833.1
4.2018.5 17 32.3 - 11375.0 2.98 19635.6
5.4019.9 27 35.0 - 16000 3.36 29324.4
9.7518.8 19 32.9 - 12333.3 3.73 23828.5
15.0019.5 44 - 442.3 221133.2 1.00 221133.2
268
Table 7-11. Soil properties from CPT2 (N) for footing 2.5 m×2.5 m
Bottom CPT (N) φ su E Eocr
(m) (bl/ft) (o) (kPa) (kPa)OCR
(kPa)GS = 0.76
1.2514 31.2 - 9500.0 8.62 27888.5
3.0019 32.7 - 11928.6 5.45 27840.0
6.5025 34.6 - 15000.0 3.44 27819.0
8.7517 32.1 - 11000.0 2.27 16569.9
12.0033 - 300.0 45000.0 1.00 45000.0
Table 7-12. Soil properties from CPT2 for footing 2.5m×2.5m
Bottom qc φ su E Eocr
(m) (bar) (o) (kPa) (kPa)OCR
(kPa)GS = 0.76
1.2541.23 39.8 - 12370.0 8.62 36313.7
3.0075.23 40.5 - 22568.6 5.45 52672.7
5.00115.94 41.1 - 34781.3 3.78 67622.3
7.7574.66 38.1 - 22399.1 2.77 37258.6
9.5049.74 35.6 - 14922.9 1.68 19355.8
12.0092.50 - 625.0 50875.0 1.00 50875.0
Modeling with CSANDSET and PLAXIS
The geometry of both footings matches with the input requirements for
CSANDSET, the main input parameters are listed in Tables 7-13 and 7-14 for Footing
3.0 m×3.0 m and Footing 2.5 m×2.5 m, respectively. The main difference in the two
footings was the number of sub-layers above the rigid layer for the Oweis (1979) method
and the Schmertmann (1978) method. Only three layers were input for Footing 3.0
269
m×3.0 m and six layers for Footing 2.5 m×2.5 m; moreover, CPT5 was combined with
SPT2 and SP3 for Footing 2.5 m×2.5 m and CPT2 with SPT2 and SPT5.
Table 7-13. Typical input for CSANDSET on footing 3.0 m×3.0 m
Input Parameter Name Input Values
Footing size (ft) 9.84
Embedment (ft) 2.49
Blow count (blow/ft) 18.8
CPT end bearing (tsf) 97.2
Total unit weight (pcf) 117.7
Saturated unit weight (pcf) 120
Coefficient K0 at rest 1.08 (DMT3)
Depth of rigid layer (ft) 36.0
Ground water table (ft) 16.0
Table 7-14. Typical input for CSANDSET on footing 2.5m×2.5m
Input Parameter Name Input Values
Footing size (ft) 8.20
Embedment (ft) 2.49
Blow count (blow/ft) 17.5
CPT end bearing (tsf) 71.9
Total unit weight (pcf) 117.7
Saturated unit weight (pcf) 120
Coefficient K0 at rest 1.65 (DMT3)
Depth of rigid layer (ft) 36.0
Ground water table (ft) 16.0
The finite element analysis with PLAXIS is achieved with axisymmetric model of
the soil-footing profile. As the actual shapes of the footings are square, thus a circular
footing with equivalent area was used; this resulted in a diameter of 3.39 m for Footing
270
3.0 m×3.0 m and a diameter of 2.82 m for Footing 2.5 m×2.5 m. The region is divided
into 15-noded triangular finite elements; the zero friction between the soil and the sides
of the footing is applied Rf = 0.01. No large deformations were expected so the
calculation stages were performed under Plastic type. The actual load steps were also
simulated in the finite element modeling with PLAXIS. A typical finite element meshing
of the domain is presented in Figure 7-37.
Figure 7-37. Axisymmetric modeling with PLAXIS of Texas A&M footings (Unit: m).
Results and Discussion
Footing 3.0 m×3.0 m
The predicted settlements from the different insitu tests with different calculation
methods are presented in Figures 7-38 through 7-43. Overall, the results obtained show
good consistency vis-à-vis the insitu data and the calculation methods. The reliability of
271
the correctly conducted (for instance: energy efficiency measurements for SPT, ASTM
standards, qualified operators) insitu tests was reflected in the results.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-38. Settlements of Footing 3.0 m×3.0 m using SPT2 data.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-39. Settlements of Footing 3.0 m×3.0 m using SPT3 data.
272
0
20
40
60
80
100
120
140
160
180
200
0 100 200 300 400 500 600 700 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1943)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-40. Settlements of Footing 3.0 m×3.0 m using CPT5 (N) data.
0
20
40
60
80
100
120
140
160
180
200
0 100 200 300 400 500 600 700 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Schmertmann (1978)
Plaxis MC
Plaxis HS
Figure 7-41. Settlements of Footing 3.0 m×3.0 m using CPT data.
273
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)Measured
DMT Method (DMT1)
DMT Method (DMT3)
Plaxis MC
Plaxis HS
Figure 7-42. Settlements of Footing 3.0 m×3.0 m using DMT data.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
PMT Method (PMT1)
PMT Method (PMT2)
Plaxis MC (PMT1)
Plaxis MC (PMT2)
Plaxis HS (PMT1)
Plaxis HS (PMT2)
Figure 7-43. Settlements of Footing 3.0 m×3.0 m using PMT data.
274
The actual settlements of the footings were more accurately predicted than those
in the Green Cove Springs footing. Not only were larger soil deformations achieved, but
also no anomalies were encountered (as opposed to Green Cove Springs adjacent
warehouse buildings, etc.). Generally, the traditional methods by far produced more
accurate settlement predictions than the finite element method with PLAXIS. From the
CSANDSET based prediction methods, D’Appolonia (1968), Shultz & Sherif (1973,
Peck & Bazaraa (1969) and NAVFAC DM 7.1 (1982) methods were the most accurate.
The family of more accurate methods were within 25% of the actual settlement at the
stress level 700 kPa. The Elastic Theory, Schmertmann (1978), Oweis (1979) and
Terzaghi (1948) methods significantly overpredicted the settlements by more than a
factor of 2. The smallest values of settlement, among the most conservative methods,
were from the Elastic Theory, at the maximum applied pressure of 800 kPa where the
actual settlement for Footing 3.0 m×3.0 m is 52.9 mm while the Elastic Theory produced
105.2 mm settlement using the SPT5 data (Figure 7-39). The traditional methods using
the DMT data and the PMT data also accurately predicted the settlements (Figures 7-42
and 7-43). The PMT method predicted the overall smallest settlement values with a
settlement of 31.0 mm at the 800 kPa of applied pressure (PMT2), but was the closest to
the actual settlement, 15.2 mm and 11.5 mm, respectively, at the 400 kPa applied
pressure.
Using SPT N based data for finite element method with PLAXIS resulted in poor
settlement predictions. The constitutive models Mohr-Coulomb and Hardening Soil,
replicate the nonlinear shape of the load-deformation curves from the load tests, but
significantly overestimate the settlements. The predicted settlements using the CPT5 (N)
275
produced the smallest values but was still larger than the values from Terzaghi (1948)
method or Elastic Theory; at the 800 kPa stress level. Both the Mohr-Coulomb model
and the Hardening Soil model produced 130.0 mm of predicted settlement versus 52.9
mm of actual settlement. Nevertheless, the finite element method appears to work well
with the CPT data, DMT data and the PMT data. The results from using the DMT data
were the most accurate followed by those from the CPT data. The rather high values of
the friction angle is the most plausible explanation for the good accuracy predictions.
Moreover, considering the good results from the DMT traditional methods, the DMT
provides a good estimation of the constrained modulus M of the sand layers. As
encountered in the Green Cove Springs footing, the Mohr-Coulomb model generally also
gives smaller predicted settlements than the Hardening Soil model for the Texas A&M
footings.
Regarding the insitu test data, it appears that the DMT provides the best data for
settlement predictions of spread footings on sand. Both the traditional methods and the
finite elements methods give reliable settlement predictions. The PMT is also reliable
although the data consistency is less, based on the discrepancy observed between the
results from the sounding PMT1 and PMT2. This could be attributed to the difference in
the amount of the soil disturbance from the two soundings and site variability.
Disregarding the Schmertmann (1978) method, the CPT is also a good insitu test for
predicting settlements of shallow footings. As for the SPT, the traditional methods
should be preferred instead of the finite element method when dealing with settlement
predictions.
276
Footing 2.5 m×2.5 m
The predicted settlements for Footing 2.5 m×2.5 m are presented in Figures 7-44
through 7-49. The same observations as those in Footing 3.0 m×3.0 m were encountered.
The actual load-settlement curve of Footing 2.5 m×2.5 m demonstrates that the
square footing is approaching its ultimate bearing capacity for applied pressures over 800
kPa. The ultimate bearing capacity of Footing 3.0 m×3.0 m was estimated to be between
the range of 1165 kPa (Terzaghi (1943)) and 1568 kPa (Schmertmann (1978)), Class
Report, Fall 1999. The effect of the smaller size of Footing 2.5 m×2.5 m only reduces
the discrepancy with the conservative methods at large deformations, on one hand, and
increases the discrepancy with the underpredicting methods at large deformations, on the
other. Nevertheless, the statements made for Footing 3.0 m×3.0 m apply to Footing
2.5 m×2.5 m as well, with different magnitudes of the settlements.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-44. Settlements of Footing 2.5 m×2.5 m using SPT2 data.
277
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-45. Settlements of Footing 2.5 m×2.5 m using SPT5 data.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Elastic Theory
D'Appolonia (1968)
Oweis (1979)
Schultz & Sherif (1973)
Terzaghi (1948)
Peck & Bazaraa (1969)
Meyerhof (1965)
NAVFAC-DM7.1 (1982)
Plaxis MC
Plaxis HS
Figure 7-46. Settlements of Footing 2.5 m×2.5 m using CPT2 (N) data.
278
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
Schmertmann (1978)
Plaxis MC
Plaxis HS
Figure 7-47. Settlements of Footing 2.5 m×2.5 m using CPT data.
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
DMT Method (DMT1)
DMT Method (DMT3)
Plaxis MC
Plaxis HS
Figure 7-48. Settlements of Footing 2.5 m×2.5 m using DMT data.
279
0
20
40
60
80
100
120
140
160
180
200
0 200 400 600 800
Applied Pressure (kPa)
Sett
lem
ent (
mm
)
Measured
PMT Method (PMT1)
PMT Method (PMT2)
Plaxis MC (PMT1)
Plaxis MC (PMT2)
Plaxis HS (PMT1)
Plaxis HS (PMT2)
Figure 7-49. Settlements of Footing 2.5 m×2.5 m using PMT data.
Conclusions
The conclusions obtained from the Texas A&M footings indeed confirmed the
results in the Green Cove Springs footing; also, some points could be made clearer. The
good quality insitu test data and suitability of the test site were considered to be
contributing factors for the better analysis of the settlement predictions. The following
conclusions can be stated as result of this study:
The estimation of the OCR values using the CPT data alone (proposed method)
can be used to predict settlement of spread footings on overconsolidated sands. The
method proposed in this study requires the use of the coefficient of lateral stress K0
obtained from Eq. (3.31). The resulting OCR values are slightly smaller than those from
Kulhawy and Mayne (1982) with the DMT K0 values, but represent a good intermediate
value for those from the other methods. The variation along the depth is also more
280
realistic compared to the Baldi et al. (1986) method. The results on the settlement
predictions showed that the CPT data and PMT data can be combined with the OCR
values to produce reasonably accurate settlement predictions.
The traditional methods, both from CSANDSET (except for CPT-based methods)
and from the insitu DMT and PMT methods generally provide more accurate settlement
predictions than the finite element method with PLAXIS. So far, the methods such as
D’Appolonia (1968), Schultz & Sherif (1973), NAVFAC DM 7.1 (1982), Peck &
Bazaraa (1969) and Meyerhof (1965) methods can be recommended over the finite
element analysis using the SPT N based soil parameters. The traditional methods such as
Elastic Theory, Terzaghi (1948), Oweis (1979) methods can be as overly too conservative
as the finite element method with PLAXIS when using the SPT N based soil properties.
The finite element method with PLAXIS produced better results with the CPT,
DMT, and PMT input parameters than with SPT and CPT (N). The high friction angle
values from the CPT data, the constrained modulus and rather high friction angle values
from the DMT data, and the secant modulus estimated from the PMT modulus with the
curve-fitted friction angles from the PMT data are considered to be the factors for the
better predictions from PLAXIS.
The Mohr-Coulomb model would be recommended over the Hardening Soil
model for predicting settlement of shallow footing on overconsolidated sands. The
evaluation of the elastic modulus of the overconsolidated sand as the normally
consolidated times the square root of the OCR value for Mohr Coulomb model appeared
to produce better results than just inputting the OCR values in the initial conditions for
the Hardening Soil model.
281
The DMT remained the insitu test that best suited the data requirements for
predicting settlement of shallow footings on overconsolidated sands. Apart from the
Self-boring PMT, based on this study, the DMT could be the second test that best
evaluates the coefficient of lateral stress K0 of the soil. The CPT also provides a good
evaluation of the initial state of stress parameters of the soil. Thus, either the DMT or the
CPT or both tests should be performed the initial stress state.
Power Plant Mat Foundation in Utah
Introduction
The Power Plant mat foundation in Utah deals with the foundations of the
Intermountain Generating Station constructed in 1981, in west central Utah. The whole
project consists of construction of two boiler buildings (the largest and heaviest
structures), two turbine generator buildings, one control building, one air control
building, and a chimney. The settlement pedictions in this study were focused on the mat
foundation of the chimney structure. The Intermountain Generating Station project was
an opportunity for the geotechnical researchers Konstantinidis et al. (1986) to conduct
comprehensive geotechnical investigations on the compressibility of overconsolidated
sands. Thus, monitoring of the actual field settlements on the site was decided and
settlement predictions based on laboratory test data and insitu test data were undertaken
for the analysis.
Site Description
The main Intermountain Generating Station structures were constructed on
gravelly sand engineered fill, compacted by vibratory rollers to dry density at least 95%
of the maximum densities. The 15 to 19 ft thick granular fill is underlain by very dense
sands and hard silt and clay layers. The dense sand layer originated from fluvial and
282
lacustrine deposits of late Tertiary and Quaternary age (Konstantinidis et al., 1986). Only
the chimney structure is directly supported by the natural soil. Figure 7-50 illustrates the
structure foundations relative to the granular fill and the native soil deposits.
chimney(to be predicted)boiler buildings
319ft
140ft
24ft
5 to
9ft 13
ft
natural deposit soils
granular engineered fill
Figure 7-50. Mat foundations relative to granular fill and natural deposits.
The chimney is founded on a reinforced concrete circular mat foundation that is 140 ft in
diameter and 13 ft thick. The mat foundation was designed to support a total weight of
the chimney superstructure 56,000kips.
Various insitu tests and conventional laboratory tests were performed for
geotechnical investigations; the insitu tests are the SPT, CPT, PMT (Self-boring and high
pressure), and seismic refraction tests. Soundings from the SPT and the CPT are in the
work of Konstantinidis et al. (1986). According to Konstantinidis et al. (1986), the sand
and clay deposits are overconsolidated due to erosion of previously overburden layers.
The data reduction for the CPT sounding is given in Appendix E, Table E-4.
283
Settlement Monitoring for Chimney Structure
Monitoring of the settlements was conducted for the boiler buildings and the
gerator buldings, as well as the chimney structure. Survey leveling techniques of high
accuracy (precision of ±0.01 in.) were used to measure the settlements. Five brass hubs
were installed around the perimeter of the chimney mat foundation to measure the
settlements. The measurement operation was started only after the mat foundations were
constructed. Thus, the chimney structure settlements represent only the settlements due
to chimney structure, whose weight is estimated as 56,000kips, an equivalent pressure of
1.819 tsf to the soil through the mat foundation; the pressure caused by the weight of the
mat foundation alone is about 0.975 tsf.
Modeling and Predictions
The geometry of the circular mat foundation of the chimney was introduced in
Figure 7-50 above. For the CSANDSET input parameters, an equivalent square footing
of 124.05 ft was used; there is no depth of embedment as the footing is lying directly on
the soil natural deposit sands. Only the SPT N values were used to derive the input
parameters for CSANDSET, the CPT (N) values could not be estimated because of the
high values of the tip resistance in the dense sands where the ratio between qc and N60 is
equal to 1 (very stiff fine grained soils), see Appendix E, Table E-4. The main input
parameters for CSANDSET are listed in Table 7-15. There are 9 sub-layers over the
depth B (width of the square footing) for Oweis (1979) method and Schmertmann (1978)
method. The water table was located around the depth 40 ft below the ground surface.
The coefficient of lateral stress K0 of the first sublayer is 1.10 and was a default value in
the main input parameters.
284
Table 7-15. Main CSANDSET inputs for chimney mat foundation
Input Parameter Name Input Values
Footing size (ft) 124.0
Embedment (ft) 0.0
Blow count (blow/ft) 66.4
CPT end bearing (tsf) 235.4
Total unit weight (pcf) 120
Saturated unit weight (pcf) 120
Coefficient K0 at rest 1.10
Depth of rigid layer (ft) 124.0
Ground water table (ft) 40.0
The finite element modeling with PLAXIS is accomplished using an axisymmetric
model, Figure 7-51. The large diameter of the footing extended the width of the finite
element half domain to 328.0 ft and the height to 130 ft, where an incompressible layer
was assumed below the end of boring from the SPT and CPT data; the bottom layer from
the sounding is identified as a very dense sand of SPT N = 83 and CPT tip resistance
around qc = 400 tsf.
Figure 7-51. Axisymmetric finite element model chimney mat foundation (Unit: m).
285
The PLAXIS input parameters from the SPT data and CPT data are given in Tables
7-16 and 7-17, respectively. The correlations listed in Chapter 6 were used.
Table 7-16. Input parameters from SPT data
Bottom SPT N γ φ su E Eocr
(ft) (bl/ft) (kN/m3) (o) (kPa) (kPa)OCR
(kPa)GS = 0.0
16.50 50 19.0 40.2 27500.0 7.18 73687.7
26.50 64 19.0 42.6 34500.0 1.48 42018.3
33.50 50 19.0 40.2 27500.0 1.83 37235.2
41.50 100 19.0 47.1 52500.0 1.75 69451.0
51.50 95 19.0 46.6 50000.0 1.65 64226.2
58.25 30 19.0 35.7 17500.0 1 17500.0
70.75 85 19.0 - 710.5 355262.7 1 355262.7
85.75 40 19.0 - 412.9 206467.3 1 206467.3
130.00 84 19.0 45.4 44500.0 1 44500.0
Table 7-17. Input parameters from CPT data
Bottom qc φ su E Eocr
(ft) (bar) (o) (kPa) (kPa)OCR
(kPa)GS = 013.94 89.96 41.5 - 26989.4 7.22 72500.9
35.27341.87 44.6 - 102560.8 1.71 133980.5
40.1995.90 37.6 - 28770.0 1.68 37298.9
48.39249.00 41.8 - 74961.0 1.83 101349.9
59.88436.26 44.1 - 130877.1 1.32 130877.1
75.4659.85 - 319.4 32916.1 1.00 32916.1
82.84172.14 39.3 - 51643.3 1.47 51643.3
85.3055.33 - 477.5 30433.3 1.00 30433.3
130.00291.60 41.5 - 87480.0 1.44 87480.0
286
Results and Discussion
The measured settlements were nearly uniform; the settlements results obtained
from five monitoring stations varied only between 0.36 in. and 0.4 in.. These are the
settlements due to the weight of the chimney superstructure alone. Two types of
prediction calculations were carried out: one from the net weight of the chimney
superstructure, and another from the difference between the settlements caused by the
mat foundation alone and the settlements caused by the whole structure (mat foundation
plus chimney superstructure). The second type of calculation is considered to be closer
to the actual conditions of the loading process. The resulting predictions are presented in
Table 7-18.
Table 7-18. Predicted settlements at Intermountain Generating Station
Structure involved Mat(1)
Total(2)
ChimneyAlone
Difference(2) – (1)
Mat Pressure (tsf) 0.975 2.794 1.819 1.819Mat Pressure (kPa) 93.4 267.6 174.2 174.2
MeasuredSettlements (in.) - - 0.36 to 0.40
Traditional MethodsTerzaghi (1948) 0.32 0.91 0.60 0.59Teng 0.23 0.65 0.42 0.42Elastic Theory 2.24 6.37 4.15 4.13D'Appolonia (1970) 0.92 2.61 1.70 1.69Peck and Bazaraa (1969) 0.20 0.58 0.38 0.38Schmertmann (1978) 1.25 3.79 2.40 2.54Schultz & Sherif (1973) 0.19 0.54 0.35 0.35Meyerhof (1965) 0.28 0.81 0.53 0.53Peck, Hanson, Thornburn 0.29 0.83 0.54 0.54Bowles 0.19 0.55 0.36 0.36NAVFAC DM 7.1 (1982) 0.17 0.49 0.32 0.32Oweis (1979) 0.47 4.00 1.65 3.53
FEM with ΠλαξισSPT 1.21 3.50 2.27 2.29
Mohr-Coulomb CPT 0.94 2.70 1.75 1.76SPT 0.43 1.23 0.81 0.80
Hardening SoilCPT 0.40 1.15 0.75 0.75
287
The results show that most of calculation methods practically did not make any
large difference between the settlements caused by the chimney alone and the settlements
caused by the whole structure minus that caused by the mat. Theoretically, given that the
past maximum stress (preconsolidation stress cσ ) has not been reached during the
loading process, the soil is assumed to behave closely to a linear elastic material and the
results from the two types of calculations should be at least approximately the same. The
exceptions were from Schmertmann (1978) method and Oweis (1979) method.
Settlement calculation in Schmertmann (1978) method evaluates the strain influence
factor (usually denoted Iz) over the depth of influence 2B for a square footing (L/B = 1).
The influence factor is a function of the geometry of the footing and the magnitude of the
footing pressure. Thus, the inconsistency of the results from Schmertmann (1978)
method could be attributed to the evaluation of the influence factors at different footing
pressure levels. Oweis (1979) method also divides the soil into layers over a depth at
least 2B or to an incompressible layer below the base of the footing, it evaluates the
modulus of deformations of each layer as dependent on the mean effective stress and the
strain level. The settlements from Oweis (1979) method show that the decrease in
modulus is more important in higher footing pressure level than it is in lower pressure
level (nonlinear variation); the result of this difference in the present case produced
settlements about twice as larger in the two types of calculations.
The most accurate settlement predictions among the traditional methods was Peck
& Bazaraa (1969) method with 0.38 in., followed by Schultz & Sherif (1973) method,
0.35 in., and the NAVFAC DM 7.1 (1982) method, 0.32 in.. D’Appolonia (1970)
288
method yielded unexpected results; D’Appolonia (1970) method overpredicted the
settlements about 4 times larger. Looking at the statistical analysis of the cases of 71
footings studied by Jeyapalan and Boehm (1986) and the FHWA Report on Spread
Footings for Highway Bridges (1987), the reason of the settlement overprediction using
D’Appolonia (1970) method coud not be explained, except the large dimensions of the
footing and the thickness of the sand layer.
The finite element method with PLAXIS overpredicted the settlements of the mat
foundation. Both the Mohr-Coulomb model and the Hardening Soil model produced
practically the same results from the two types of calculation. In the Hardening Soil
model, the preconsolidation stress cσ was not reached and the hardening cap was not
expanded; so the soil deformation was purely elastic and the settlements in the two
calculation types were almost identical. In contrast to the cases of the previous spread
footings, the Hardening Soil model produced more accurate results than the Mohr-
Coulomb model. Comparing the results from the SPT data and CPT data, the Hardening
Soil model also shows better consistency than the Mohr-Coulomb model does. The
linear elastic behavior below the yield stress would lead to smaller settlement predictions
over the nonlinear hardening flow. Konstatinidis et al. (1986) reported that based on the
laboratory consolidation tests performed on the sand layers, the maximum past stress is
between 5.0 tsf to 10 tsf (over depth 100 ft); the pressure applied by the chimney
superstructure after installation of the mat foundation is 1.82 tsf; the magnitudes of the
pressures indicate that the mat foundation pressure from the chimney is rather small
compared with the preconsolidation of the soil. This observation also applies with the
Green Cove Springs footing; thus, the only explanation for the anomaly between the
289
Hardening Soil model and the Mohr-Coulomb model could be again the large dimensions
of the mat foundation and its impact on the the two different types of OCR input
parameters (square root for the Mohr-Coulomb model and POP in the initial conditions
stage for the Hardening Soil model).
Conclusions
Few conclusions could be drawn from the case of the Power Plant mat foundation
in Utah. The special case of the present settlement prediction was the large dimension of
the footing, which implies larger bearing capacity; it is aggravated by the very dense sand
layers (average SPT N = 66.4) over only the depth B below the base of the mat
foundation. Below the sand layer, no information about the soil properties is known and
the assumption of a rigid layer was adopted. Some of the conclusions are extension of
the previous conclusions to the cases of large footings or mat foundations, or limitations
of some calculations methods to the smaller spread footings only.
Even in large footings (mat foundation), the traditional methods generally provide
more reliable settlement predictions than the finite element method with PLAXIS. The
more accurate methods were Schultz & Sherif (1973), Peck & Bazaraa (1969), NAVFAC
DM 7.1 (1982), and Meyerhof (1965). D’Appolonia (1970) method, until more cases of
mat foundations are studied, may not suit the settlement prediciton in mat foundation on
relatively shallow sand layers. The Elastic Theory, Terzaghi (1948), Schmertmann
(1978), and Oweis (1979) methods remain overly conservative in the settlement
prediction of the mat foundation.
The finite element constitutive models also overpredicted the settlements even in
the case of the mat foundations. However, the Hardening Soil model appeared to better
handle the overconsolidated behavior of the sand layer than the Mohr-Coulomb model.
290
The predicted settlement using the Hardening Soil model was only twice as much as the
measured ones (100% larger), which can be useful for a design purpose. The Mohr-
Coulomb model gave poor results from both the SPT dat and the CPT data.
The insitu data from SPT and CPT data are useful data when combined with the
traditional methods using the correlations selected in this study. They could also be used
conservatively with the Hardening Soil model of PLAXIS but more similar cases should be
studied for validation of the statement.
Conclusion for Shallow Footings
As general conclusion for the settlements of shallow footings, few points can be
made. The insitu tests SPT, CPT, DMT, and PMT are useful for predicting the footing
settlements. That statement considers that the appropriate method was used for the insitu
test. The SPT, although best used with the traditional method has to be carefully
selected. The data from the Texas A&M University project were the type of good quality
data for the SPT: energy efficiencies were measured, test conducted correctly under
ASTM standard, etc. The PMT, when used with the PMT method of settlement
calculation is accurate as well. But as was observed, the soil disturbance during the
boring has to be minimized. The type of PMT data in the Texas A&M University project
is tolerably good. Different methods were used by the operators to avoid excessive
disturbance. The CPT and the DMT were the best tests in predicting settlements of
shallow footings in this study. Data from these two insitu tests are as useful for
traditional methods as for the finite element method with PLAXIS, excluding the
Schmertmann (1978) method. Thus, they are strongly recommended. Moreover, with
the absence of the Self-boring PMT, which is more rarely used, the CPT and DMT are
excellent tools to evaluate the stress history of overconsolidated sands. Consistency of
291
the data was also very convincing. Two data sets are obtained from the CPT: the
equivalent blow count CPT (N) data (good for the traditional method) and the actual CPT
data (good for the finite element method). The author suggests that CPT using CPT (N)
should be selected over the SPT. The DMT is mandatory for a best estimation of the
initial state of stress of sand layers. The DMT data provided the least variation of results
from the various methods in this study.
The following traditional methods are recommended (but not systematically) for
calculations of settlements of shallow footings on sands: D’Appolonia (1968), Schultz &
Sherif (1973), Meyerhof (1965), DMT method, PMT method. The finite element method
with PLAXIS is useful only when combined with the CPT data and the DMT data. And
for that, the Mohr-Coulomb model should be preferred over the Hardening model and the
Soft Soil model.
292
CHAPTER 8STATNAMIC LOAD TESTS ON SHALLOW FOUNDATIONS
General
This chapter deals with Statnamic load testing on shallow foundations, the
Statnamic analyses using the conventional method with SAW R4, and the dynamic finite
element method with PLAXIS. Supplementary details on the Statnamic load testing are
presented in addition to the introduction in Chapter 2. Two Statnamic load tests on
spread footings were performed and analysed in this study: the concrete footing in Green
Cove Springs (footing of Chapter 6), and a steel plate footing in Orlando. Unlike the
types of calculations in the static load tests in Chapter 6 and Chapter 7 where the load-
settlement curves were predicted using various calculation methods, the actual Statnamic
load-displacement results from the Statnamic load tests are first evaluated using the
spreadsheet SAW R4. Then, the corresponding static load-settlement curves and the
damping coefficients are derived using the Unloading Point Method in the same
spreadsheet. The finite element analysis will consist of the assessment of the input
parameters in dynamic type of calculations by simulating mainly the actual load-
displacement curves from the Statnamic load test.
Statnamic Load Testing
The Statnamic load test on shallow footing represents an alternative to the tedious
and time-consuming static load test. The rapid development of the Statnamic load test
devices, such as the catching systems, the reaction mass types, encourages more
engineers and researchers to select the Statnamic load test over the static load test. The
293
fundamental idea of the Statnamic load test is based on the Newton’s second law and the
Newton’s third law. The resultant of the external forces acting on an object is equal to
the product of its mass and its acceleration in magnitude and in direction, the vertical
Statnamic force applied to the foundation structure as a result of the combustion process
produces acceleration which will depend on the total mass of the displacing objects. The
reciprocity of forces on two objects also applies in that the reaction mass, when pushed
upward by the high pressure from the combustion chamber will cause a downward force
applied to the foundation structure. A Statnamic test would accelerate the reaction
masses from at rest to around 20g’s in less than 0.1s (Lewis, 1999).
Equipment
At present, few types of Statnamic load test devices are in use, the latest type is
the hydraulic catching system with the capacity up to 4.0 MN, the largest to date
(Mullins, 1998). Other types of catching system are the gravel catching system
associated with reaction masses composed of annular vessels filled with gravel; and the
mechanical catching system, which is used only on submerged reaction mass system.
Details on the last two types of catching system can be found in Lewis’ Master thesis
(1999). Only the mechanism of the Statnamic load test with the hydraulic catching
system that was used in Green Cove Springs and in Orlando is discussed in this section.
There are five main components in the Statnamic load test equipment: the piston,
the silencer, the hydraulic catching system, the reaction mass, and the venting apparatus.
The reaction masses for the hydraulic catching system are donut-shaped made of
reinforced concrete and covered with thin wall steel shell; hollow steel cans are also
being used, which is more practical for transportation. The different parts of the
Statnamic test equipment fully set up are sketched in Figure 8-1.
294
Footing
Piston
Combustionchamber Steel plate
Silencer
Venting system
Hydraulic catching mechanicsm
Reaction mass
Load cell
Accelerometers
Figure 8-1. Statnamic equipments set up for a load test (lines not shown).
295
As explained in details by Lewis (1999), the Statnamic device is working
similarly to an ordinary internal combustion engine. The only difference is that the
piston, which also plays the role of the combustion chamber, remains stationary and
launches the reaction masses upward when high pressure results from the burning of the
fuel nitrocellulose (mixture of nitroglycerin and paper). The release of the pressure
(highly pressured gases) takes place when the “silencer”, which also carries the reaction
masses on its reaction flange, is fully detached from the piston. The silencer and the
reaction masses are prevented from returning by the four hydraulic actuators in the
catching mechanism; the four hydraulic rams, which act independently (for redundancy
and safety reasons) are activated by four low pressure (1500 psi) nitrogen gas on top of
hydraulic oil, when the silencer and the reaction masses take off, the nitrogen gas
expands and let the hydraulic oil fill the four nitrogen accumulators through the one-way
valves from the oil chamber; with the oil being unable to flow back, the silencer and the
reaction masses are retained floating in the air (Mullins, 1998). For a subsequent
Statnamic load test cycle, the fuel basket will only need to be recharged in the piston,
then the hydraulic oil is redirected into lower vessel, which lowers the silencer and the
mass reactions to the test-ready position. Illustration of the two positions of the silencer
and the reaction masses is provided in Figure 8-2. This process enables performance of
up to 20 Statnamic cycles on multiple foundations in one day for a well-trained crew
(Lewis, 1999).
Test Setup and Procedure
Apart from the five main components of the Statnamic equipment,
instrumentation of the structure foundations and data acquisition are also of great
importance in a Statnamic load tests.
296
Hydraulic actuators (or rams)
A B
Figure 8-2. Positions of silencer and reaction masses. A) Before test. B) After test
The other components are the load transducer, the displacement transducer (either
an accelerometer or a laser/photovoltaic sensor), and the data acquisition with
MEGADAC computer system. The load transducer consists of a strain gage-based load
cell with accuracy within 0.1% of the actual Statnamic load; the load cell is mounted on
the piston, Figure 8-1. The laser/photovoltaic sensor has an accuracy of 0.001 mm and is
also mounted on the piston; the laser ray is sent to a photovoltaic sensor through a laser
window, which establishes a reference datum (Lewis, 1999). The second type of
displacement transducer is the acclerometer; it measures directly the magnitude of the
acceleration of the foundation and enables calculation by integration of the velocity and
the displacement versus time. The accelerometers were used for the statmamic load test
performed in Green Cove Springs and in Orlando. The sketch of the Statnamic test setup
297
in Figure 8-1 shows the diametrally-opposed locations of the two accelerometers on top
of the shallow concrete footing. The data acquisition system MEGADAC is equipped
with software that can monitor almost any displacement and load trasducers. Figure 8-3
displays the fully hooked up data acquisition system that was used in Green Cove
Springs.
Figure 8-3. Data acquisition with MEGADAC system in Green Cove Springs.
The process begins with the preparation of the footing, circular concrete footing
in Green Cove Springs and circular steel plate in Orlando. For the concrete footing, a
circular steel plate is placed on top of the footing; the levelled flat surface to
accommodate the piston and distribute uniformly the load is mandatory in Statnamic
tests. The piston is tied with the plate with screw and bolts (Figure 8-1). The functioning
of the measuring devices is also checked at this stage. The structural catching system is
then installed followed by the placement of the silencer from the top of the catching
system with a crane. The reaction masses are also lowered the same way as the silencer.
298
Before proceeding to the actual test, all necessary checks should be carried out, such
checks include the ignition circuits, leaks and damages in the various lines (hydraulic and
electric), the data acquisition, functioning of measuring devices, all of Statnamic devices
well centered with respect to the frame of the hydraulic catching system, etc.
The actual Statnamic test starts with filling the basket fuel in the piston as shown
in Figure 8-4, then lowering the silencer and the reaction masses to close the combusion
chamber cylinder formed by the piston and the silencer.
A B
Figure 8-4. Process before actual Statnamic test. A) Filling fuel basket. B) Loweringsilencer and reaction masses.
When the piston-silencer connection is sealed, the ignition circuit is started and the
Statnamic load applied resulting the combustion, Figure 8-5.
299
A B
Figure 8-5. Actual Statnamic test process. A) Before test (circuit ignition). B) during test.
The data are transferred into the MEGADAC data acquisition system in form of files
*.00X, where X in the extension is the number of the Statnamic cycle performed. The
*.00X files can be opened with ASCII format files for further anlaysis and data reduction.
Statnamic Test on Green Cove Springs Footing
The Statnamic load test on the Green Cove Springs footing was a part of the insitu
test and numerical modeling project. The test was performed on February 24, 2002,
about eight months after the static load test was terminated. The static part of Green
Cove Springs was achieved in Chapter 6 and the dynamic part is the object of this
section. Comparison between the Statnamic derived static results and the results in
Chapter 6 is also made in this section.
300
Maximum Statnamic Loads at Green Cove Springs
Five Statnamic load cycles were performed on the Green Cove Springs footing in
the early afternoon of February 24, 2002. The whole test lasted about three hours
including the setting up of the different line connections (hydraulic and electric). The
maximum Statnamic loads reached in each cycle are listed in Table 8-1.
Table 8-1. Maximum Statnamic loads on Green Cove Springs footing
Cycle Maximum Load(kN)
Pressure(tsf)
Cycle 1 502 2.0
Cycle 2 698 2.8
Cycle 3 1231 4.9
Cycle 4 2411 9.6
Cycle 5 3842 15.3
Cycle 1 could not be used in the analysis due to unsatisfactory quality of the data; also,
Cycle 5 was exluded from the analysis because the pressure 15.3 tsf (Table 8-1) was
considered to be out of range for the type of analysis in this study. Thus, only Cycle 2,
Cycle 3 and Cycle 4 are analyzed throughout the Statnamic analysis for the Green Cove
Springs footing.
Preliminary Parametric Study
The idea of the soil mass that is in-phase with the spread footing when displaced
by the Statnamic force was already discussed in Chapter 3. The quantity of the moved
soil mass was determined by doing a parameteric study from different points of view.
The first assumption was that the displaced soil mass is delimited by the wedge from the
bearing capacity theory; that is the portion of the foundation soil in passive state. The
301
bearing capacity theories are plane strain-based theory (strip footing), such as that of
Terzaghi (1943) and Rankine (1857). The second assumption is the result of the
discussion with Dr. Townsend, University of Florida, admitting that the soil mass moved
by the dynamic load is determined from the law of pressure distribution down to depth B,
which is also the width of the footing. Sketches of the in-phase soil wedges displaced by
Statnamic force are provided in Figure 8-6. The common point in the two assumptions is
that the quantity of the soil mass depends on the size of the footing; the main difference is
that in the former, the volume of the soil mass is dependent on the friction angle of the
soil whereas in the latter the slope of the wegde is constant for any soil (2 to 1 or 30°
slope). In the Terzaghi (1943) bearing capacity theory, the angle between the base of the
footing and the wedge is equal to the value of the friction angle of the soil φ; in the
theories such as Rankine (1857), Meyerhof (1951), and Hansen (1970), that angle is
equal to (45° + φ/2).
α
30°or
Terzaghi (1943) α = φ Hansen (1970), Rankine (1857) α = 45° + φ/2
In-phase soil mass
Base of footingBB
B
21
Figure 8-6. In-phase soil mass during a Statnamic test.
302
A parametric study on the volume of the soil mass was conducted to see how
much it affects the dynamic parameter results: the damping coefficient, and the derived
maximum static force. Five cases were examined using the spreadsheet SAW R4:
1. Case 1 uses the bearing capacity theory with the Terzaghi (1943) assumption (α = φ,Figure 8-6) and using the smallest value of the friction angle of the soil, which wasobtained from the SPT data: φ = 31.4°;
2. Case 2 uses the bearing capacity theories with the assumption α = 45° + φ/2, andusing the largest value of the friction angle, which was obtained from the DMT data:φ = 46.3°;
3. Case 3 uses the bearing capacity theories with the Terzaghi (1943) method but usingthe average value of the friction angles from all of the insitu test data (SPT, CPT,DMT, and PMT): φ = 39.0°;
4. Case 4 uses the bearing capacity theories with the assumption α = 45° + φ/2, butusing the average value of the friction angles from all of the insitu test data (SPT,CPT, CPT (N), DMT, and PMT): φ = 39.0°;
5. Case 5 uses the second assumption of pressure distribution but rather than 30.0° ofinclination the 2 to 1 slope is used, which produces less volume of displaced soil, tobe compared to the other four cases.
Table 8-2. Parametric study on Cycle 2
CASE 1 2 3 4 5 AverageUnit weight (kN/m3) 19.0 19.0 19.0 19.0 19.0 -Friction angle (o) 31.4 46.3 39.0 39.0 39.0 -Volume of soil (m3) 0.49 2.00 0.65 1.68 11.21 -Mass of soil (kg) 946.5 3867.2 1255.7 3251.1 21709.5 -Maximal load (kN) 688.5 688.5 688.5 688.5 688.5 -Statnamic pressure (tsf) 2.74 2.74 2.74 2.74 2.74 -
OutputAverage damping (kNs/m) 1007.7 1003.4 1007.26 1004.3 976.9 1005.7Median damping (kNs/m) 1089.9 1067.5 1088.08 1069.81 1020.4 1078.8Static load (kN) 678.2 681.05 678.49 680.5 698.53 679.56Static pressure (tsf) 2.70 2.71 2.70 2.71 2.78 2.70
303
Table 8-3. Parametric study on Cycle 3
CASE 1 2 3 4 5 AverageUnit weight (kN/m3) 19.0 19.0 19.0 19.0 19.0 -Friction angle (o) 31.4 46.3 39.0 39.0 39.0 -Volume of soil (m3) 0.49 2.00 0.65 1.68 11.21 -Mass of soil (kg) 946.5 3867.2 1255.7 3251.1 21709.5 -Maximal load (kN) 1242.7 1242.7 1242.7 1242.7 1242.7 -Statnamic pressure (tsf) 4.95 4.95 4.95 4.95 4.95 -
OutputAverage damping (kNs/m) 2013.9 2211.4 2034.8 2169.8 3418.0 2107.5Median damping (kNs/m) 2043.6 2295.4 2070.3 2242.3 3724.0 2162.9Static load (kN) 1170.9 1180.9 1172.0 1178.8 1242.5 1175.6Static pressure (tsf) 4.66 4.70 4.66 4.69 4.94 4.68
Table 8-4. Parametric study on Cycle 4
CASE 1 2 3 4 5 AverageUnit weight (kN/m3) 19.0 19.0 19.0 19.0 19.0 -Friction angle (o) 31.4 46.3 39.0 39.0 39.0 -Volume of soil (m3) 0.49 2.00 0.65 1.68 11.21 -Mass of soil (kg) 946.5 3867.2 1255.7 3251.1 21709.5 -Maximal load (kN) 2428.5 2428.5 2428.5 2428.5 2428.5 -Statnamic pressure (tsf) 9.66 9.66 9.66 9.66 9.66 -
OutputAverage damping (kNs/m) 1939.1 2048.7 1950.7 2025.6 2718.2 1991.0Median damping (kNs/m) 1806.1 1935.3 1819.8 1908.0 2667.1 1867.3Static load (kN) 1797.1 1829.0 1800.5 1822.3 2023.8 1812.2Static pressure (tsf) 7.15 7.28 7.16 7.25 8.05 7.21
Tables 8-2 through 8-4 summarize the parameters resulting from the different
cases. Generally, the assumption of the 2 to 1 pressure distribution of Case 5 produced
the largest values in the estimation of damping coefficient and the maximum derived
static load. In Cycle 2, the results from the different cases are more uniform; the
discrepancy becomes more pronounced in Cycle 3 and Cycle 4.
While the damping coefficients and the derived maximum static load remained
close within Case 1 through Case 4, the damping coefficient in Case 5 increased to about
304
1.5 times larger than the other cases (Cycle 2). Plot of the derived static load versus
settlement demonstrated a very mediocre and atypical form of such curve. Within Case 1
through Case 4, all of the assumptions appear to provide reasonable results for the
Statnamic analysis; Case 4 shows the most consistent results and is always closest to the
average values from Cycle 2 to Cycle 4. It is decided that the Statnamic analysis in this
study is undertaken with the assumption of Case 4: bearing capacity theory by Hansen
(1970) and others and average friction angle from the various insitu test data.
Conventional Analysis
Using the Unloading Point Method, the damping coefficients in Cycle 2, Cycle 3
and Cycle 4 are evaluated; the Statnamic load-settlement curves in the three cycles are
established as well; subsequently, the corresponding static load-settlement curves are
derived. Listed in Table 8-5 are the mass inputs lumped together for the calculations.
Table 8-5. Equivalent mass of Green Cove Springs footing
Structure Unit Mass
Concrete footing (6ft diam.) [kg] 3842.5
Piston [kg] 500.0
Steel plate [kg] 1590.0
In-phase soil mass [kg] 3251.1
Load-Settlement curves
The Statnamic pressure-settlement curves along with the derived static pressure-
settlement cuves are plotted in Figure 8-7.
305
0.0
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1.0
0 1 2 3 4 5 6 7 8 9 10Applied Pressure (tsf)
Sett
lem
ent (
in)
Statnamic Derived Static
Figure 8-7. Pressure-displacement curves: Cycle 2, Cycle 3, and Cycle 4.
The soil damping coefficients corresponding to each cycle have been given in
Tables 8-2 through 8-4.
Comparison with static load test results
By picking up the maximum values of the derived static pressure in the plots of
Figure 8-7, the comparison with the results from the static load test in Chapter 6 is
achieved. Such comparison plot is presented in Figure 8-8. Cycle 4 was trimmed off the
comparison plot as the maximum static pressure was 2.3 tsf; on the other hand, the
derived static peak pressure in Cycle 1 and its corresponding settlement were added to the
plot. It is observed that despite the bad quality of the data from Cycle 1, the choice of
only picking the peak value for the comparison was reasonable in that linear elastic
behavior of the pressure-settlement was reproduced at the early stage of the loading. For
comparison with the Unloading Point Method, the prediction methods that were selected
from the analysis in Chapter 6 were the most accurate methods from each insitu test data.
306
0.0
0.2
0.4
0.6
0.8
1.0
1.2
1.4
0 1 2 3 4 5
Applied Pressure (tsf)
Sett
lem
ent (
in)
Measured
D'Appolonia (1968)
Schultz & Sherif (1973)
Peck & Bazaraa (1969)
DMT Method
PMT Method
Plaxis MC (CPT)
Plaxis MC (DMT)
Derived Static (STN)
Cycle 4
Figure 8-8. Comparison of true static and Statnamic derived static results.
Discussion
The plots in Figure 8-8 show that the Statnamic derived static curve best matches
the measured settlements from the static load test in Green Cove Springs. Densification
of the soil may have occured during the static load test, which was performed about eight
months before the Statnamic load test. But given the overconsolidation state of the sand
layer as discussed in Chapter 6, such factor cannot be accounted for the rather smaller
settlement from the Statnamic derived static load-settlement curve. For instance, it was
estimated from the Kulhawy and Mayne (1982) in Table 6-4 that the overconsolidation
ratio of the sand at depth 2.5ft below the ground surface is OCR = 28.1; the past
maximum stress at depth 2.5ft would then be around 4.2 tsf, which is much smaller than
the effective stress increase induced by the surface pressure 2.3 tsf at the base of the
fooitng during the static load test. Thus, the stress path during the static loading was
located inside the elastic region in the case of Hardening Soil model, and no further
307
overconsolidation of the sand has taken place. Good agreement of the results from the
actual static load test and the Statnamic load test are not uncommon in practice; examples
of such cases have been achieved by Matsumoto (1998), and Mullins (1998).
Finite Element Modeling with PLAXIS
Finite element simulation of the dynamic behavior of the soil-structure system
with Πλαξισ consists of examining the validity of the dynamic input parameters. The
basic input parameter is the dynamic shear modulus G0, which also can be evaluated from
the shear wave velocity Vs, Eq. (2.31). Correlations between the SPT N value or CPT
properties and the dynamic properties of soil have been developed by various researchers,
as those mentioned in Eq. (3.20) and Eq. (3.37), respectively. For simplicity, the
correlation of Eq. (3.20) was selected to estimate the shear wave velocity of a sand layer.
It has produced satisfactory results in the Statnamic finite element analyses on piles by
Horikoshi et al. (1998). The using of correlation Eq. (3.20) requires the multiplying of
the initial shear modulus G0 to a reduction factor η (η ≤ 1). The sand layer in Green
Cove Springs being of Pleistocene age (Chapter 6), the basic input parameters were
evaluated from the equivalent blow count CPT (N) in Table 8-6. The axisymmetric
model is now extended to a much larger half-domain: 30.0 m wide and 15.0 m high for
the 1.82 m diameter and 0.61 m thick concrete footing. The absorbent boundaries were
imposed at the bottom boundary and the right side boundary to avoid any reflected waves
from the rigid fixities; the meshing discretized the domain into 15-noded triangular
elements. The meashing is generally coarse except in the regions surrrounding the
footing where it is more refined by factor of 4. Figure 8-9 illustrates the description of
the finite element model with PLAXIS.
308
Table 8-6. Basic soil properties for dynamic finite element modeling
Bottom γ CPT (N) φ or su Vs G0 E0
(m) (kN/m3) (bl/ft) (o) or (kPa) (m/s) (kPa) (kPa)GS = 0
2.0019.0 24.5 35.8 273.13 143459.9 372995.7
3.0017.5 14.5 32.4 230.56 93709.7 243645.1
4.2518.0 18.2 33.7 248.12 113462.4 295002.2
6.0016.0 9.5 30.4 201.13 64827.2 168550.8
10.0015.6 3.0 38.5* 163.53 42427.9 110312.7
*undrained cohesion
Figure 8-9. Dynamic numerical modeling of Green Cove Springs footing (Unit: m).
As the concrete footing was also simulated in the geometry, only the weights of the
piston and the steel plate were taken into account in the traction load A-A of Figure 8-9.
The pressure caused by the piston and the steel plate through the footing was about 7.8
kN/m2. The calculation process is executed in two phases: the first phase is to assign the
309
properties of the elastic concrete to the footing, which was set as soil cluster in the initial
conditions; the second phase is to enter the value 7.8 kN/m2 and the Statnamic pulse load
file, the latter is taken from an ASCII format file. Finally, the viscosity parameters:
Rayleigh coefficients α and β are entered at the calculation stage.
The constitutive model that is most suitable to simulate the dynamic behavior of
the soil is the Linear Elastic model.
Discussion on the coefficients η, α and β
The reduction factor η is multiplied to the initial shear modulus of the soil for
modeling the dynamic elastic behavior of the soil. It has a direct impact on the load-
settlement of the footing. In the Green Cove Springs footing, a change of the value of η
from 1 to 0.5 (in other words, the shear modulus from the initial value G0 to its half)
would increase the maximum settlement from 1.0 mm to 2.2 mm in Cycle 2 with an
assumption of a viscous soil foundation. The residual settlement was also increased by
almost twice as much (0.3 mm to 0.6 mm) by the same change of the factor η. Specific
values of η are also required according to the choice of the constitutive models in order to
obtain the same results. For the same problem, the value of η in Mohr-Coulomb model
would generally be larger than that in Linear Elastic model. For instance, the η = 0.3 for
Linear Elastic model and η = 0.4 for Mohr-Coulomb model in Cycle 4 would produce
close displacement-time plot.
The Rayleigh coefficients α and β are arbitrary coefficients used to determine the
damping matrix by proportionality to the mass and the stiffness of the soil (Zienkiewicz,
1991; Clough and Penzien, 1975). According to PLAXIS manual, a greater value of α
than β would damp more of the low frequency vibrations whereas greater values of β
310
than α would damp more of the high frequency vibrations. The experience in solving the
Green Cove Springs footing showed that the Rayleigh coefficient α did not cause major
effect on the displacement-time curve. The larger the value of the coefficient β, the
bigger the residual displacement is, the smaller the peak displacement, and the longer the
time to reach the peak displacement. For the simulation of the Statnamic load test on the
Green Cove Springs footing, the values of α were generally greater than the values of β.
Results and discussion
Plot of the Statnamic applied pressure, and the vertical displacements versus time
are given in Figure 8-10, for Cycle 2, Cycle 3 and Cycle 4. Calculations using the soil
properties from the SPT data were also attempted. For the Linear Elastic model and the
same values of Rayleigh damping coefficients, the peak vertical displacements and the
residual displacements were generally larger than those from the CPT (N). Combination
of the plots in Figure 8-10 was then utilized to establish the pressure-settlement curves
and their finite element simulation with PLAXIS in Figure 8-11.
The plots of Figure 8-10 exhibit good agreement between the peak values of the
displacements. However, the simulated residual displacements could not be well
matched with the actual values, exact estimation of the damping coefficients α and β
could be one of the reasons of the mismatch. It was observed that in Cycle 4, the footing
settlement kept increasing even if the Statnamic load was no more in action (almost zero
at time t = 0.3s), Figure 8-10; such phenomenon could not be characterized in the
simulation and is hard to explain in the real case, except the imperfections in the
measuring devices (accelerometers).
311
0
100
200
300
400
500
600
700
800
900
1000
0.0 0.1 0.2 0.3 0.4 0.5
Time (s)
Stat
nam
ic P
ress
ure
(kPa
)
-0.025
-0.023
-0.020
-0.018
-0.015
-0.013
-0.010
-0.008
-0.005
-0.003
0.000
0.0 0.1 0.2 0.3 0.4 0.5
Time (s)
Def
lect
ion
(m)
Cycle 2: η = 0.375; α = 0.2; β = 0.005 – Cycle 3: η = 0.300; α = 0.2; β = 0.005Cycle 4: η = 0.300; α = 0.2; β = 0.005
Figure 8-10. Comparison of measured and numerical simulation with PLAXIS.
312
0.0
0.1
0.2
0.3
0.4
0.5
0.6
0.7
0.8
0.9
1.0
0 1 2 3 4 5 6 7 8 9 10Applied Pressure (tsf)
Sett
lem
ent (
in)
Measured Calculated with Plaxis
Figure 8-11. Finite element simulation of Pressure-Settlement with PLAXIS.
The plots of Figure 8-11 show a good estimation of the elastic modulus of the
soils, both in loading and in unloading cases. The values of the reduction factor η for the
shear modulus varied between 0.300 and 0.375. The gap between the displacements
could result from the different estimation of the damping coefficients, as mentioned
above. Nevertheless, the Rayleigh damping coefficients α = 0.2 and β = 0.05 (Figure 8-
10) appear to be good choices in simulation of the viscosity of the soil in Green Cove
Springs.
Conclusions
Some useful conclusions could be drawn from the analysis of the Statnamic load
test performed on the Green Cove Springs footing. The Statnamic load test offers a good
alternative solution to the static load test in estimating the compressibility of the soil. In
addition to being much less tedious, less time-consuming and more economical,
313
comparison of the actual static load test settlements with the Statnamic derived static load
settlements was successful.
Based on the shape of the load displacement curves and consistency of the results,
it is more reasonable to obtain the volume of the in-phase soil mass from the bearing
capacity theory rather than from the 2 to 1, or 30° pressure distribution theory. The
change in friction angle values does not affect significantly the peak derived static load
and the damping coefficients.
In the finite element analysis with PLAXIS, the assumption of Linear Elastic
behavior of soils for dynamic loading is more appropriate than Mohr-Coulomb model or
Hardening Soil model in simulating the dynamic behavior of the soil.
The correlation between the shear wave velocity and the CPT (N) blow count, Eq.
(2.31), is a useful and practical tool for estimating the dynamic modulus of the soil.
Knowledge of the geological age of the sand or clay layer proved critical; combination of
the aforementioned equation with the reduction factor around 0.300 to 0.375 also
appeared reasonable to evaluate the shear modulus input in the finite element analysis.
The Rayleigh damping coefficients were slightly harder to estimate in order to
best match the residual displacement of the footing. However, reasonably good order of
the arbitrary values of α and β could be used in modeling the viscosity of the sands at
Green Cove Springs.
Orlando Steel Plate Footing
Introduction
The Orlando Statnamic load test was a part of the ASCE –GI Deep Foundation
Congress that took place in Orlando on October 12, 2001. Statnamic load test on a steel
314
plate of diameter 54in was one of the test exhibitions at the site in Orange County, Figure
8-12. The main purpose was to demonstrate how practical and fast Statnamic tests can be
conducted. A number of insitu tests were also carried out for soil investigations: SPT,
CPT, and DMT.
Figure 8-12. ASCE-GI Deep Foundation Congress 2001 test site.
Orlando Test Site
The Orange County outcrop and shallow subcrop layers are of Pleistocene-
Holocene age in the Eastern parts and Miocene in the western parts. The County is also
known for producing natural materials such as crushed stones, sands, and peats
(Randazzo and Jones, 1997). The insitu tests performed for the Statnamic load tests
showed loose to medium dense sands down to a depth of 50ft, underlain by very soft
clays. Based on verbal information communicated verbally, the location of the Statnamic
315
test on the steel plate footing was on the boundary separating the Pleistocene and the
Miocene areas south of the Orlando International Airport.
Two CPT were performed by the Ardaman & Associates, Inc., but only one data
set was used for the analysis; the location of the CPT sounding from the Statnamic load
test on the steel plate was not clearly known. The DMT was performed by the contractor
Wright-Padgett-Christopher; the DMT thrust values were not recorded during the test.
No information about the location of the DMT sounding with respect to the steel plate
footing was available either. The SPT was also performed but apparently the data were
not reliable. Thus, the analysis of the Statnamic load test was limited to the uses of the
data from one CPT and one DMT (for traditional methods) only. The soil profile based
on the sounding CPT1 is presented in Figure 8-13; while that based on the DMT
sounding is in Figure 8-14.
Statnamic Load Test in Orlando
Only one cycle of Statnamic load test was conducted on the steel plate footing at
the Orlando site. The maximum Statnamic load test reached was around 710 kN at a time
of 0.2s. Unfortunately, the data collected did not appear to be complete as the decreasing
value part of the load with time was missing from the data. A plot of the load versus
integrated displacement (from the accelerometer) also showed a nonregular Statnamic
loading of the steel plate. Due to inconsistencies and poor quality of the Statnamic data,
dynamic analysis of the test could not be achieved. Using the Unloading Point Method,
the ultimate derived static load was evaluated, as 604 kN; and the corresponding vertical
displacement was 6.23 mm. This data point was the only usable data from the Statnamic
load test on steel plate in Orlando.
316
Figure 8-13. CPT sounding in Orlando test site (Ardaman & Associates, 2001).
317
0.0
0.5
1.0
1.5
2.0
2.5
3.0
3.5
4.0
4.5
0 2 4 6 8 10
Material Index I D
Dep
th (m
)
0.0
0.5
1.0
1.5
2.0
2.5
3.0
3.5
4.0
4.5
0 1000 2000 3000 4000
Constrained Modulus M (bar)
Dep
th (m
)
0.0
0.5
1.0
1.5
2.0
2.5
3.0
3.5
4.0
4.5
0.0 0.2 0.4 0.6 0.8 1.0
Undrained Shear Strength s u (bar)
Dep
th (m
)
0.0
0.5
1.0
1.5
2.0
2.5
3.0
3.5
4.0
4.5
0 10 20 30 40 50
Horizontal Stress Index K D
Dep
th (m
)
Figure 8-14. DMT sounding in Orlando test site, Wright-Padgett-Christopher.
318
The analysis then consists of predicting the settlement 6.23 mm from the applied
pressure of 405.88kPa or 4.26 tsf (derived static load of 604 kN) on the circular steel
plate of 54in in diameter and 6in thick.
Modeling with CSANDSET and PLAXIS
The soil properties from the CPT sounding were used for the soil stratigraphy,
strength and stiffness parameters of the layers. The DMT soil properties, due to lack of
data, were used only in the settlement calculation with the DMT method. Soil layering
using the equivalent blow count from the CPT data were also established; Tables 8-7 and
8-8 show the soil properties from the CPT and CPT (N), respectively. The properties in
Table 8-8 were combined with the input parameters for CSANDSET that are listed in
Table 8-9. The exception is that the coefficient of lateral stress at rest K0 was determined
from the CPT data rather than from the DMT data.
Table 8-7. Soil properties from CPT data
Bottom γ qc φ E Eocr
(m) (kN/m3) (bar) (o) (kPa)OCR
(kPa)GS = 0
3.50 18.0 54.8 39.8 16427.3 14.5 62507.0
5.34 19.0 195.6 44.9 58665.0 3.1 103937.2
15.50 18.5 81.3 38.3 24402.2 3.0 41931.0
Table 8-8. Soil properties from CPT (N) data
Bottom γ CPT (N) φ E Eocr
(m) (kN/m3) (bl/ft) (o) (kPa)OCR
(kPa)GS = 0
3.50 18.0 16.2 32.1 15608.7 14.5 59392.0
5.34 19.0 38.6 38.2 26800.0 3.1 47481.7
15.50 18.5 24.1 34.5 19559.7 3.0 33610.0
319
Table 8-9. Input parameters for CSANDSET
Soil Properties over depthB = 4.0ft below Base Input Values
Blow count (bl/ft) 16.2
CPT end bearing (tsf) 57.21
Total unit weight (pcf) 110.0
Saturated unit weight (pcf) 114.5
Coefficient K0 at rest 2.0 (CPT)
Depth of rigid layer (ft) 50.0
Ground water table (ft) 2.0
Results and Discussion
The measured settlement in the Orlando project was taken as the endpoint of the
derived static load-settlement curve from the Statnamic load as mentioned above. Thus,
the measured settlement itself does not represent the actual settlement if an actual static
load test was to be performed on the Orlando site. The very good agreement between the
actual settlement and the derived static settlement in the case of Green Cove Springs
footing was taken as confirmation that the measured settlement could be replaced with
the Statnamic derived static settlement. Comparison with the results from the various
methods of settlement predictions is presented in Table 8-10.
The overestimation of the settlement from the various methods occurred again for
the Orlando steel plate footing. The most accurate prediction was that from the DMT
method: 0.32in, which is within 30% of the measured settlement 0.24in; the good
settlement prediction using the DMT method was also encountered in the Green Cove
Springs and Texas A&M cases.
320
Table 8-10. Predicted settlements of Orlando steel plate footing
Derived static Load (kN) 604.11Pressure (kPa) 405.88Pressure (tsf) 4.26
Settlement (inches)Measured (Derived Static) 0.24
InsituDMT Method 0.32
CSANDSETTerzaghi (1948) 3.68Elastic Theory 1.24D'Appolonia (1970) 0.44Peck and Bazaraa (1969) 0.50Schmertmann (1978) 2.14Schultz & Sherif (1973) 0.59Meyerhof (1965) 0.91NAVFAC DM 7.1(1982) 0.93Oweis (1979) 1.42
PLAXIS
Mohr-Coulomb 0.42Hardening Soil 1.54
Among the traditional methods, D’Appolonia (1970) method came closest with
0.44in but is almost twice as large as the measured settlement; Peck & Bazaraa (1969)
and Schultz & Sherif follow with settlements 0.50in and 0.59 in, respectively. These
three methods have performed well too in the previous three cases especially when used
with CPT (N) data. The methods such as Meyerhof (1965) and NAVFAC DM 7 (1982)
are slightly more conservative than the previous results. The rest of the methods: Oweis
(1979), Terzaghi (1943), and Schmertmann (1978) were overly conservative which is not
an exception based on the results from the previous studies.
The finite element method with Mohr-Coulomb model also produced a good
settlement prediction. The CPT data, which is characterized by the high friction angle
values, appeared to be more suitable with the Mohr-Coulomb model than with the
Hardening Soil model. The predicted settlements were rather distant, 0.42in and 1.54in
321
athough both models included the OCR values in the calculations but in different ways.
It is the stress level dependent stiffness, as opposed to the constante stiffness in Mohr-
Coulomb that is considered to be responsible for the excessive settlement estimated with
the Hardening Soil model. In fact, except the case of Power Generating Station in Utah,
the Hardening Soil model generally produced larger settlements. The combination Mohr-
Coulomb model and CPT data appear to give good results in the Orlando settlement
predictions.
The discrepancy between the measured and the predicted settlements can be
attributed to the inaccuracy of the finite element method with PLAXIS. In this particular
case, the equivalent blow count CPT (N) data were useless as no settlement predictions
could be obtained. The predictions from the actual CPT data were yet conservative.
Conclusion
Despite the incompleteness of the data obtained from the Orlando Statnamic load
test, several confirmation conclusions could be drawn from the analysis.
The insitu test CPT has been proven to be a very useful test in estimating the
different soil properties and using them for further analysis both with the traditional
methods and the finite element method with PLAXIS, especially the Mohr-Culomb model.
The equivalent blow counts CPT (N) are key parameters for SPT based traditional
methods. The DMT is also highly recommended for settlement analysis of shallow
footings on sands. The constrained modulus estimated from the DMT represents well the
elastic property of the sand layers. Even without the DMT Thrust data, which prevents
DMT based finite element analysis, settlement prediction with traditional method (DMT
method) is still recommended.
322
Among the traditional methods, settlement predictions using D’Appolonia (1970)
method, Schultz & Sherif (1973) method, and Peck & Bazaraa (1969) remained among
the most reliable whereas Schmertmann (1978) method, Oweis (1979), and Terzaghi
(1943) are still overly conservative; these are results from the CPT (N) based soil
properties.
The Mohr-Coulomb model in the finite element method with PLAXIS can produce
good prediction when using the CPT data. The Hardening Soil model in the Orlando case
resulted in an excessive value of the predicted settlement. It appears that higher values of
the strength properties and stiffness properties could help improve the accuracy of the
predictions in the finite element method. However, correlations providing such
estimations have not been developed and therefore may be of interest. For instance, the
OCR value from the Dilatometer data reduction program, Bullock (1983), or
Schmertmann based on Kulhawy and Mayne (1982), which was derived from calibration
chamber test sands, may be extended to cases of silty sands and sandy silts, etc. soils for
finite element analyses with PLAXIS. The possible drawback is that it can be less
conservative analysis.
Conclusion for Statnamic Load Tests
The Statnamic test was proven to be a good alternative to static load test for
shallow footings on sands. It is considerably less time consuming than the static load
test: six days of static test can be performed in half day with Statnamic load test (five
cycles). The key method for the derived static load-settlement results is the Unloading
Point Method. The Unloading Point Method yielded good results for the cases of the
shallow footings at Green Cove Springs and Orlando.
323
The CPT is a good test for the soil properties and thus still recommended. The
equivalent blow count CPT (N) proved effective to evaluate the soil properties for the
conventional method and the finite element method with PLAXIS. The SPT was not as
good as the CPT in both Green Cove Springs and Orlando. The data were not of good
quality and the analysis with SPT data-derived soil properties for the Green Cove Springs
footing in the finite element method was etremely tedious yet unsuccessful.
The finite element modeling of the Statnamic load test with PLAXIS is feasible.
And the most appropriate constitutive model for the type of problem (Statnamic load test
for shallow footing on sands) is the Linear Elastic model. Estimation of the elastic
properties of the sands and the Rayleigh damping coefficients could be achieved with
reasonable accuracy.
324
CHAPTER 9CONCLUSIONS AND RECOMMENDATIONS
In this research, application of insitu test data to differents types of geotechnical
problems has been achieved. Four insitu tests were selected: the SPT, the CPT, the
DMT, and the PMT; and three categories of geotechnical problems: the unloading cases
in sheet pile walls, the loading case in shallow footings, and the dynamic case in
Statnamic load test. The study consisted of a total of eleven calculations using traditional
methods and numerical method with PLAXIS. The calculations were limited to the small
deformation of the soil and the structures only. Effects of creep and long-term behavior
were not included. Also, rather than obtaining the most accurate predictions (deflections)
for the studied cases, a more conservative approach and at the same time more consistent
procedure was undertaken by setting a certain factor of safety between the measured
deflections and the predicted ones.
Results from this study are: conversion of the PMT modulus into triaxial modulus
was proposed; evaluation of the friction angle using the curve-fitting method with PLAXIS
also yielded acceptable results; the using of the unload-reload modulus in the Mohr-
Coulomb model was very effective; methods for the determination of the stress history of
sand layers were developed; and finite element modeling of Statnamic load tests on
shallow footings were successfully achieved. The following conclusions could be drawn
from this study.
325
Sheet Pile Wall Problems
It is unfortunate that the DMT was never performed in any of the five sheet pile
wall projects, so no conclusion on the test could be obtained. The SPT, the PMT were
more or less equally useful in the predictions of sheet pile wall deflections. The
suefuleness of the insitu tests along with the methods and the constitutive models is
summarized in Table 9-1.
Table 9-1. Summary for sheet pile wall problems
FEMR: RecommendedNR: Not Recommended Traditional
Mohr-Coulomb Hardening Soil
SPT NR R only with EUR a) b) R a) b)
CPT (N) NR R only with EUR b) R b)
CPT NR R b) c) R
DMT — — —
PMT NR R only with EUR a) R a)
a) Testing procedure criticalb) Not for soft clayc) For a more conservative analysis
Shallow Footing Problems
For shallow footing problems, the four insitu tests all were useful in predicting
settlements. Testing procedures in SPT, DMT and PMT are critical in obtaining accurate
and reliable predicted settlements. Table 9-2 summarizes the usefulness of the
combinations of the insitu tests and the settlement calculation methods.
326
Table 9-2. Summary for shallow footing problems
FEMR: RecommendedNR: Not Recommended Traditional
Mohr-Coulomb Hardening Soil
SPT R a) b) NR NR
CPT (N) R a) NR NR
CPT NR c) R d) e) R d)
DMT R R e) f) R f)
PMT R b) NR NR
a) Not: Oweis (1979) method, Terzaghi (1948) method, Elastic Theoryb) Testing procedure criticalc) Schmertmann (1978) methodd) For conservative analysise) Preferable over Hardening Soil modelf) DMT Thrust critical
Statnamic Problems
The Statnamic derived static load-settlement curve accurately matched the actual
static load test on the Green Cove Springs footing. Based on the results mentioned
previously, the Unloading Point Method resulted in settlements smaller than those from
the traditional methods and finite element methods. The smaller settlement was also
encountered in the case of the Orlando footing. In this study, the Unloading Point
Method provided a closer simulation of the load-settlement curve of the actual static load
test than did the dfferent traditional methods and finite element method with PLAXIS in
static cases.
For the Statnamic load test on shallow footings on sands, the soil is most suitably
modeled as a Linear Elastic material, and the viscosity, which is represented by the
damping matrix, can be determined using the Rayleigh approach, as proportional to the
stiffness matrix and the inertia matrix.
327
The CPT can be used for dynamic modeling with finite element method. In this
study, the equivalent blow count CPT (N) proved to be good data for estimating the
dynamic stiffness of the linear elastic soil when combined with the reduction factor η.
The advantages of performing the Statnamic load test over the actual static load
test were evident for shallow footings on sands. The Statnamic load test was much less
time consuming and less tedious, half a day for 5 cycles versus 5 days of monotonic static
load test. The Statnamic load test should become more practical with the rapid
development of the test equipment, and it represents a valid, factor of safety-wise,
alternative to the static load test.
Recommendations and Future Work
In order to better evaluate the soil properties of the soil from the insitu tests, the
following future works are also proposed:
Development of an insitu constitutive model for PMTing will help back-calculate
and refine the evaluation of the strength and stiffness properties of the soil from the PMT
pressure-deformation curves.
Better understanding of the viscosity parameters of the soil would be obtained by
developing the relationship between the Rayleigh proportionality coefficients α and β in
the damping matrix and the type (or behavior) of the soil.
328
APPENDIX ABASIC EQUATIONS OF THE DIFFERENT METHODS
Traditional Method with CWALSHT
Coulomb’s Earth Pressure Theory (Homogeneous Backfill)
The weight of the wedge (ABE) in the configuration of Figure A-1 is given
below.
( ) ( )( )
−+
×+××
=βρβαρα
αγ
sinsinsin
sin2 2
2HW (A.1)
( ) ( )δφραφρ ++−−°=
− 180sinsinWPa (A.2)
After c
δ
αρ
δ
φ
γ φ δ
β
ψ
θθ = α − δ
ψ = ρ − φ
180° − θ − ψ
Figure A-1. Coulomb’s equilibrium theory.
Combining Eq. (A.1) and Eq. (A.2) and knowing that Pa depends on angle ρ, the
maximum active wall force Pa is obtained from setting dPa/dρ = 0
( )
( ) ( ) ( )( ) ( )
22
22
sinsinsinsin1sinsin
sin2
+×−−×+
+×−×
+×
×=
βαδαβφδφδαα
φαγ HPa (A.3)
aa KHP ××
=2
2γ (A.4)
329
Sweep Search Wedge Method Slices
The equilibrium of horizontal and vertical forces presented in Figure A-2,
excluding the wall slice gives
( )θφ
θθφtantan
sectantan11 ±
×±×±×=− −
i
iiiii
CWPP (A.5)
where Pi, Pi-1, normal forces on left and right side vertical surfaceWi, weight of each slice including any surface surcharge on the sliceφi, effective friction angle at the bottom of the sliceθi, angle of inclination of the surfaceCi, effective cohesion of the soil at bottom of slice times its length
Figure A-2. Sweep search wedge method of slices.
Subsequently, equilibrium of the wall-slices yields:
( )( )
×±±×±
=
×±×±±
φθ
θφθδθφθδ
sincos
sintancoscoscostansinsin /
wn
aww
w
PA
wav
wav
CPFCW
NP
(A.6)
where( )∑
∑ ×=
j
jjav h
h δδ is the weighted average wall friction angle
( )∑ ×= jja ahF is the wall-soil adhesion force.
330
Traditional Methods with CSANDSET
The equations included are: Elastic Theory Eq. (A.7); Schmertmann Method
(1978), Eq. (A.8); D’Appolonia Method (1968), Eq. (A.12); Oweis Method (1979), Eq.
(A.13); Schultz and Sherif Method (1973), Eq. (A.14); Terzaghi Method (1948), Eq.
(A.15); Peck and Bazaraa Method (1969), Eq. (A.16); and Meyerhof Method ((1965), Eq.
(A.17).
Fss
IImE
vBq ×××−
×′×=2
01ρ (A.7)
where q0, the contact pressure in units of Es
B’, least lateral dimension of the contributing area in the units of ρIs and IF, influence factors depending on L’, B’, Poisson’s Ratio ν,embedment D, and H the thickness of stratumEs, elastic modulus of the soil
∑= ×
∆××′×××=
n
i cii
izi
qxHI
qCCC1
321ρ (A.8)
′′
−=q
C Dσ5.011 (A.9)
+=
1.0log2.012
tC (A.10)
BLC 03.003.13 −= (A.11)
where C1, depth factorC2, secondary creep factorC3, shape factorσ’D, effective stress at base of footing, depth Dn, number of soil layersIzi, influence factor at midpoint of soil layer I∆Hi, thickness of layer Ixi, factor by which to multiply the CPT end bearing qci to obtain Young’smodulus Eit, time since application of contact pressure q’ (in years)B and L, footing dimensions
331
10 µµρ ×××
=M
Bq (A.12)
where q, footing bearing pressureB, footing widthM, modulusus of compressibility of sandµ0, embedment correction factor (depending on D and B)µ1, correction factor for thickness of sand layer
( )11
−=
−××
= ∑ ii
n
i iFF
EBqρ (A.13)
where n, number of layers of soilq, net bearing pressure at footing levelB, footing widthEi, equivalent liner soil modulus for layer iFi-1, settlement factor at top of layer IFi, settlement factor at bottom of layer I
+××
××=
BD
BBN
fBqcm4.0171.1
)(
1
87.0
ρ (A.14)
where q, footing bearing pressure in kg/cm2
f, correction factor for footing shape and thickness of the sand stratumB, footing width in cmB1 = 1cmN, SPT blow countD, embedment of the footing
2
123.)(
+×
××=
BB
NPCCin Dwρ (A.15)
where Cw, water table correction (1.0 ≤ Cw ≤ 2.0)CD, embedment correction factor = [1 – D/(4B)]B, footing width in ftD, embedment of the footing in ftP, footing bearing pressure in tsfN, SPT blow count
332
2
122)(
+×
××=
BB
NPCCinchesB
Dwρ (A.16)
Same meaning for the variable names as in Terzaghi method except NB, SPT blow count
corrected for overburden pressure; and Cw and CD calculated differently.
2
122.)(
+×
×=
BB
NPCin Dρ (A.17)
The variables are evaluated exactly the same designation as those from the Terzaghi
method.
( )21
2
14)(
+×
×=
BKBqft
V
ρ (A.18)
where q, footing bearing pressure in tsfB, footing width in ftKV1 = modulus of vertical subgrade reaction for 1ft square bearing plate atground surface.
Settlement with Insitu Tests
The insitu method of settlement calculations includes the PMT method, Eq.
(A.19) and the DMT method, Eq. (A.22).
BqEB
BBqE c
cd
d×××+
××××= λαλρ
α
*9
*9
2
00 (A.19)
1EEc = (A.20)
++++=
16/98/7/65/4/321 5.21
5.211
85.011
411
EEEEEEd(A.21)
where Ek, equivalent pressuremeter modulusq*, net average bearing pressure, = q – q0q, average bearing pressureq0, total vertical stress at the depth of the vase of the footingB0, reference width, usually 60cm (2.0ft)B, width or diameter of the footing (B ≥ B0)α, rheological factor depending on the soil type and the ratio Ek/pl
λc, λd, shape factors, depending on the footing dimensions: L an B
333
i
n
i i
vi HM
∆×′∆
= ∑=1
σρ (A.22)
where ∆σ'vi, effective stress increase caused by the footing bearing pressure q inlayer i, (using Boussinesq or Westergraad method)Mi, tangent constrained modulus of layer i∆Hi, thickness of layer i, usually 200mm
Figure A-3. Excel Spreadsheet for calculating settlement with PMT method.
334
Statnamic Test: Unloading Point Method
Figure A-4 represents the lumped mass m of the lumped footing and in-phase soil
mass, the soil stiffness k and damping coefficient C.
Load Fstn
Time t
Funl
Displacement u
Time tuunl
Time t
v = 0; t = tumax
Acceleration a
Time taunl
Velocity vFstn
Funl
FuFstn(max)
Load Fstn
Displacem
ent u
m
C
k
Fsoil
Fstn
Figure A-4. Statnamic signals and load-displacement diagrams.
Determination of Static Resistance from Unloading Point
For u = displacement (measured) or a = acceleration (measured) and known value
of the statnamic force Fstn
dtduv = (A.23)
2
2
dtuda = (A.24)
The total reaction of the soil is
vusoil FFF += (A.25)
Fu = k×u = static resistance (unknown)Fv = C×v = damping force (unknown), where C is the damping factor;Fa = m×a = inertial force
335
The equilibrium equation is
avuasoilstn FFFFFF ++=+= (A.26)
amvCFF stnu ×−×−= (A.27)
At maximum displacement (Unloading Point)
maxmax at time 0 uttuuv ==⇔= (A.28)
( ) ( ) and maxmax uunlustnunl taatFF == (A.29)
Finally
( ) unlunluu amFtF ×−=max (A.30)
Construction of Derived Static Load-Displacement Diagram
The Unloading Point Method assumes that the soil is yielding over the range
Fstn(max) to Funl (Figure A-4), thus
unlu FF = (A.31)
Over this range, the following equations are valid
aunlstnv FFFF −−= (A.32)
vFFF
C aunlstn −−= (A.33)
then the Mean damping factor Cmean is calculated for the above range. The static
resistance Fu is finally given by
ameanstnu FvCFF −×−= (A.34)
The derived static load-displacement curve, Fu versus u, can be plotted as in Figure A-4.
336
APPENDIX BSHEET PILE AT MOFFITT CANCER CENTER
Soil Properties from CPT, Moffitt Cancer Center
The friction angle calculated in the program Cptintr1 v. 3.04, Tables B-1 and B-2
is, from the correlations by Robertson and Campanella (1983). The correlations in the
calculations are by Kulhawy and Mayne (1990).
337
Table B-1. Soil properteis from CPT test, North, at Moffitt Cancer Center
University of Florida
Operator :JBA/LR/HM CPT Date :12-04-00 13:56 On Site Loc:Moffitt Center Cone Used :156 Job No. :MF1 Water table (meters) : 4.1 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 7.55 44.90 0.59 0.02 sand to silty sand >90 >48 19 UNDEFINED 0.50 1.64 23.21 106.80 0.46 0.07 sand >90 >48 46 UNDEFINED 0.75 2.46 23.70 124.94 0.53 0.11 sand >90 >48 47 UNDEFINED 1.00 3.28 21.47 113.54 0.53 0.16 sand >90 >48 43 UNDEFINED 1.25 4.10 13.04 60.54 0.46 0.20 sand >90 >48 26 UNDEFINED 1.50 4.92 7.64 27.70 0.36 0.25 sand to silty sand 70-80 44-46 19 UNDEFINED 1.75 5.74 5.15 19.22 0.37 0.29 sand to silty sand 60-70 42-44 13 UNDEFINED 2.00 6.56 4.63 15.38 0.33 0.34 sand to silty sand 60-70 42-44 12 UNDEFINED 2.25 7.38 4.18 15.54 0.37 0.38 sand to silty sand 50-60 40-42 10 UNDEFINED 2.50 8.20 4.71 19.08 0.41 0.43 sand to silty sand 50-60 40-42 12 UNDEFINED 2.75 9.02 5.10 22.30 0.44 0.47 sand to silty sand 50-60 40-42 13 UNDEFINED 3.00 9.84 5.82 24.78 0.43 0.52 sand to silty sand 60-70 40-42 15 UNDEFINED 3.25 10.66 6.58 26.14 0.40 0.56 sand to silty sand 60-70 40-42 16 UNDEFINED 3.50 11.48 8.00 28.80 0.36 0.61 sand to silty sand 60-70 42-44 20 UNDEFINED 3.75 12.30 9.04 34.48 0.38 0.65 sand to silty sand 70-80 42-44 23 UNDEFINED 4.00 13.12 10.21 40.54 0.40 0.70 sand 70-80 42-44 20 UNDEFINED 4.25 13.94 12.44 51.72 0.42 0.74 sand 70-80 42-44 25 UNDEFINED 4.50 14.76 10.11 47.90 0.47 0.76 sand 70-80 42-44 20 UNDEFINED 4.75 15.58 6.68 142.82 2.14 0.78 silty sand to sandy silt 50-60 40-42 22 UNDEFINED 5.00 16.40 9.05 223.40 2.47 0.80 silty sand to sandy silt 60-70 40-42 30 UNDEFINED 5.25 17.22 10.56 225.50 2.14 0.82 silty sand to sandy silt 70-80 42-44 35 UNDEFINED 5.50 18.04 11.50 144.18 1.25 0.84 sand to silty sand 70-80 42-44 29 UNDEFINED 5.75 18.86 10.40 142.66 1.37 0.86 sand to silty sand 70-80 40-42 26 UNDEFINED 6.00 19.69 3.10 117.18 3.78 0.88 clayey silt to silty clay UNDFND UNDFD 15 2.9 6.25 20.51 3.65 125.88 3.45 0.90 clayey silt to silty clay UNDFND UNDFD 18 3.5 6.50 21.33 5.57 99.42 1.79 0.92 silty sand to sandy silt 50-60 38-40 19 UNDEFINED 6.75 22.15 4.48 108.06 2.41 0.94 sandy silt to clayey silt UNDFND UNDFD 18 4.3 7.00 22.97 1.93 52.40 2.71 0.97 clayey silt to silty clay UNDFND UNDFD 10 1.8-----------------------------------------------------------------------------------------------------------------------------------
338
Table B-1. continued University of Florida
Operator :JBA/LR/HM CPT Date :12-04-00 13:56 On Site Loc:Moffitt Center Cone Used :156 Job No. :MF1 Water table (meters) : 4.1 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 7.25 23.79 1.92 56.20 2.92 0.99 clayey silt to silty clay UNDFND UNDFD 10 1.7 7.50 24.61 2.15 68.64 3.19 1.01 clayey silt to silty clay UNDFND UNDFD 11 2.0 7.75 25.43 1.82 57.80 3.17 1.03 clayey silt to silty clay UNDFND UNDFD 9 1.6 8.00 26.25 1.98 56.52 2.85 1.05 clayey silt to silty clay UNDFND UNDFD 10 1.8 8.25 27.07 2.36 48.12 2.04 1.07 sandy silt to clayey silt UNDFND UNDFD 9 2.2 8.50 27.89 1.48 37.48 2.53 1.09 clayey silt to silty clay UNDFND UNDFD 7 1.3 8.75 28.71 1.01 11.92 1.18 1.11 clayey silt to silty clay UNDFND UNDFD 5 .8 9.00 29.53 1.03 12.14 1.17 1.13 clayey silt to silty clay UNDFND UNDFD 5 .8 9.25 30.35 1.55 43.32 2.80 1.15 clayey silt to silty clay UNDFND UNDFD 8 1.3 9.50 31.17 1.97 37.88 1.92 1.17 sandy silt to clayey silt UNDFND UNDFD 8 1.8 9.75 31.99 0.53 12.18 2.28 1.19 silty clay to clay UNDFND UNDFD 4 .3 10.00 32.81 1.72 39.98 2.32 1.21 clayey silt to silty clay UNDFND UNDFD 9 1.5 10.25 33.63 9.56 144.02 1.51 1.23 silty sand to sandy silt 60-70 38-40 32 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 10
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
339
Table B-2. Soil properties from CPT test, South, at Moffitt Cancer CenterUniversity of Florida
Operator :JBA/LR/HM CPT Date :12-04-00 14:42 On Site Loc:Moffitt Center Cone Used :156 Job No. :MF2 Water table (meters) : 4.1 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 12.17 72.16 0.59 0.02 sand >90 >48 24 UNDEFINED 0.50 1.64 24.30 135.34 0.56 0.07 sand >90 >48 49 UNDEFINED 0.75 2.46 23.16 168.30 0.73 0.11 sand >90 >48 46 UNDEFINED 1.00 3.28 16.60 104.94 0.63 0.16 sand >90 >48 33 UNDEFINED 1.25 4.10 10.47 66.62 0.64 0.20 sand to silty sand >90 46-48 26 UNDEFINED 1.50 4.92 6.98 41.56 0.60 0.25 sand to silty sand 70-80 44-46 17 UNDEFINED 1.75 5.74 6.12 38.42 0.63 0.29 sand to silty sand 70-80 44-46 15 UNDEFINED 2.00 6.56 5.98 37.42 0.63 0.34 sand to silty sand 60-70 42-44 15 UNDEFINED 2.25 7.38 5.84 37.34 0.64 0.38 sand to silty sand 60-70 42-44 15 UNDEFINED 2.50 8.20 6.80 42.26 0.62 0.43 sand to silty sand 60-70 42-44 17 UNDEFINED 2.75 9.02 8.86 51.80 0.58 0.47 sand to silty sand 70-80 42-44 22 UNDEFINED 3.00 9.84 10.18 57.22 0.56 0.52 sand to silty sand 70-80 42-44 25 UNDEFINED 3.25 10.66 11.83 61.06 0.52 0.56 sand 80-90 44-46 24 UNDEFINED 3.50 11.48 14.70 75.52 0.51 0.61 sand 80-90 44-46 29 UNDEFINED 3.75 12.30 16.55 88.78 0.54 0.65 sand 80-90 44-46 33 UNDEFINED 4.00 13.12 8.78 45.10 0.51 0.70 sand to silty sand 60-70 40-42 22 UNDEFINED 4.25 13.94 6.81 107.28 1.57 0.74 silty sand to sandy silt 60-70 40-42 23 UNDEFINED 4.50 14.76 6.99 182.82 2.62 0.76 sandy silt to clayey silt UNDFND UNDFD 28 6.9 4.75 15.58 4.05 204.46 5.05 0.78 clay UNDFND UNDFD 40 3.9 5.00 16.40 2.69 138.00 5.14 0.80 clay UNDFND UNDFD 27 2.5 5.25 17.22 2.50 120.20 4.80 0.82 clay UNDFND UNDFD 25 2.4 5.50 18.04 3.43 139.22 4.06 0.84 silty clay to clay UNDFND UNDFD 23 3.3 5.75 18.86 2.47 133.42 5.39 0.86 clay UNDFND UNDFD 25 2.3 6.00 19.69 2.34 112.24 4.80 0.88 clay UNDFND UNDFD 23 2.2 6.25 20.51 2.26 99.58 4.40 0.90 clay UNDFND UNDFD 23 2.1 6.50 21.33 4.91 238.96 4.86 0.92 silty clay to clay UNDFND UNDFD 33 4.7 6.75 22.15 3.09 83.72 2.71 0.94 sandy silt to clayey silt UNDFND UNDFD 12 2.9 7.00 22.97 2.91 73.92 2.54 0.97 sandy silt to clayey silt UNDFND UNDFD 12 2.7-----------------------------------------------------------------------------------------------------------------------------------
340
Table B-2. continued
University of Florida
Operator :JBA/LR/HM CPT Date :12-04-00 14:42 On Site Loc:Moffitt Center Cone Used :156 Job No. :MF2 Water table (meters) : 4.1 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 7.25 23.79 2.50 64.82 2.59 0.99 clayey silt to silty clay UNDFND UNDFD 12 2.3 7.50 24.61 1.91 50.76 2.65 1.01 clayey silt to silty clay UNDFND UNDFD 10 1.7 7.75 25.43 1.50 42.48 2.83 1.03 clayey silt to silty clay UNDFND UNDFD 7 1.3 8.00 26.25 1.39 28.54 2.05 1.05 clayey silt to silty clay UNDFND UNDFD 7 1.2 8.25 27.07 1.36 5.44 0.40 1.07 sandy silt to clayey silt UNDFND UNDFD 5 1.2 8.50 27.89 3.74 -1.68 -0.04 1.09 undefined UNDFND UNDFD UDF UNDEFINED 8.75 28.71 1.49 3.08 0.21 1.11 sandy silt to clayey silt UNDFND UNDFD 6 1.3 9.00 29.53 2.55 9.12 0.36 1.13 silty sand to sandy silt <40 32-34 8 UNDEFINED 9.25 30.35 9.90 255.26 2.58 1.15 sandy silt to clayey silt UNDFND UNDFD 40 9.7 9.50 31.17 5.46 119.36 2.19 1.17 sandy silt to clayey silt UNDFND UNDFD 22 5.2 9.75 31.99 7.46 147.90 1.98 1.19 silty sand to sandy silt 50-60 38-40 25 UNDEFINED 10.00 32.81 19.82 -12987.14 -65.52 1.21 undefined UNDFND UNDFD UDF UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 10
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
341
Data Reduction in Slope Inclinometer Test
The Digitil Inclinometer model used at Moffitt Cancer Center produces digital
readings equal to 2 times the sinθ, θ is the angle of inclination as indicated in Figure B-1.
During the reading process, the decimal point was ignored (for simplification purpose)
but will be accounted in the data reduction. Thus, the readings A+, B+, A-, and B- are
RDG = (2sinθ)×10+4 (B.1)
θi
θi
Cumulative Deflection
Read
ing
ever
y 1.
0ft
Figure B-1. Inclinometer measurement and data reduction.
(1) The difference of the initial readings A0i and A180i (before the excavation) is firstcalculated to obtain the column with header abbreviated “Init. Diff.”, (A+ = A0 andA- = A180).
Init. Diff. = RDG(A0i) – RDG(A180i) (B.2)
(2) Then, the algebraic difference of the final readings A0f and A180f (after theexcavation).
Alg.. Diff. = RDG(A0f) – RDG(A180f) (B.3)
342
(3) The CHANGE is obtained from the difference between the “Alg. Diff” and the “Init.Diff.”.
CHANGE. = Alg. Diff. – Init. Diff. (B.4)
(4) The deflection is obtained from
Spacing) Readings(000,104
CHANGESpacing) Readings(sin ××
=×=∆ θ (B.5)
The same procedure is followed to calculate the deflection in B direction.
Figure B-2. Mini-CPT and equivalent SPT N at Moffitt Cancer Center site. (fromArdaman & Associates, Inc.)
343
PMT at depth 10.0ft
-2
0
2
4
6
8
10
12
-50 0 50 100
Volume (cm3)
Pres
sure
(bar
) Insitu
Morh-Coulomb
Hardening Soil
PMT at depth 15.0ft
-2
0
2
4
6
8
10
12
-50 0 50 100
Volume (cm3)
Pres
sure
(bar
) Insitu
Morh-Coulomb
Hardening Soil
PMT at depth 20.0ft
-2
0
2
4
6
8
10
12
-20 0 20 40 60 80 100
Volume (cm3)
Pres
sure
(bar
) Insitu
Morh-Coulomb
Hardening Soil
PMT at depth 25.0ft
-2
0
2
4
6
8
10
12
-20 0 20 40 60 80 100
Volume (cm3)
Pres
sure
(bar
) Insitu
Morh-Coulomb
Hardening Soil
Figure B-3. PMT curve-fitting at Moffitt Cancer Center.
344
APPENDIX CSHEET PILE WALLS IN KARLSRUHE, HATFIELD AND ROTTERDAM
Table C-1. Results of the sheet pile wall field test at Karlsruhe site
Stage 1 Stage 2 Stage 3 Stage 4 Stage 5 Stage 6 Stage 7Depth (m) Horizontal Displacements (mm)
0.00 2.45 7.75 8.25 7.55 7.05 5.55 5.150.10 2.30 7.46 7.90 7.26 6.83 5.35 5.010.60 1.57 5.95 6.18 5.65 5.61 4.33 4.231.10 0.95 4.33 4.37 4.18 4.35 3.37 3.551.60 0.34 2.61 2.65 2.85 3.39 2.78 3.062.10 0.05 1.25 1.21 1.90 2.63 2.72 3.112.60 -0.17 0.31 0.32 1.14 2.17 2.88 3.363.10 -0.17 0.05 0.04 0.62 1.72 2.99 3.403.60 -0.17 -0.05 -0.04 0.19 1.11 2.67 3.114.10 -0.17 -0.10 0.00 0.00 0.45 2.14 2.434.60 0.11 0.05 0.09 0.14 0.20 1.49 1.655.10 0.11 0.05 0.09 0.09 0.10 0.64 0.785.60 0.00 0.00 0.00 0.00 0.00 0.00 0.006.00 0.00 0.00 0.00 0.00 0.00 0.00 0.00
Depth (m) Bending Moments (kNm/m)0.00 0.00 0.00 0.00 0.00 0.00 0.00 0.001.00 -0.89 -1.25 -2.66 -2.55 -3.58 -4.41 -5.062.00 -0.55 -2.26 -2.12 0.26 1.37 -1.53 1.703.00 -0.08 -0.75 -0.93 -1.14 1.14 2.20 2.764.00 -0.05 -0.03 0.02 -0.56 -1.56 0.37 0.785.00 -0.02 -0.02 0.06 0.11 -0.09 -1.18 -1.706.00 0.00 0.00 0.00 0.00 0.00 0.00 0.00
Depth (m) Strut Forces (kN/m)1.25 - - 4.29 13.88 21.31 28.64 33.72
(Source: Wolffersdorff, 1997)
345
Figure C-1. CPT test soundings at Karlsruhe site. (Source: Wolffersdorff, 1997)
346
Table C-2. Soil properties from sounding CPT D at Karlsruhe site
University of Florida Operator :e Wall CPT Date :etten Sheet Pil On Site Loc:Karlsruhe Univ. Cone Used :156 Job No. :CPT D Water table (meters) : 5.5 Tot. Unit Wt. (avg) : 16.5 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 3.56 30.00 0.84 0.02 silty sand to sandy silt >90 >48 12 UNDEFINED 0.50 1.64 5.92 55.40 0.94 0.06 silty sand to sandy silt >90 >48 20 UNDEFINED 0.75 2.46 3.72 29.60 0.80 0.10 silty sand to sandy silt 70-80 46-48 12 UNDEFINED 1.00 3.28 3.56 28.57 0.80 0.14 silty sand to sandy silt 60-70 44-46 12 UNDEFINED 1.25 4.10 5.71 50.00 0.88 0.19 silty sand to sandy silt 70-80 44-46 19 UNDEFINED 1.50 4.92 8.20 71.43 0.87 0.23 sand to silty sand 80-90 46-48 20 UNDEFINED 1.75 5.74 7.74 68.95 0.89 0.27 sand to silty sand 70-80 44-46 19 UNDEFINED 2.00 6.56 5.84 29.95 0.51 0.31 sand to silty sand 60-70 42-44 15 UNDEFINED 2.25 7.38 10.56 80.20 0.76 0.35 sand to silty sand 80-90 44-46 26 UNDEFINED 2.50 8.20 9.76 83.25 0.85 0.39 sand to silty sand 80-90 44-46 24 UNDEFINED 2.75 9.02 11.56 95.75 0.83 0.43 sand to silty sand 80-90 44-46 29 UNDEFINED 3.00 9.84 15.74 124.00 0.79 0.47 sand >90 46-48 31 UNDEFINED 3.25 10.66 11.23 86.90 0.77 0.52 sand to silty sand 80-90 44-46 28 UNDEFINED 3.50 11.48 10.46 84.90 0.81 0.56 sand to silty sand 70-80 42-44 26 UNDEFINED 3.75 12.30 9.47 80.40 0.85 0.60 sand to silty sand 70-80 42-44 24 UNDEFINED 4.00 13.12 6.78 67.00 0.99 0.64 sand to silty sand 60-70 40-42 17 UNDEFINED 4.25 13.94 7.55 68.10 0.90 0.68 sand to silty sand 60-70 40-42 19 UNDEFINED 4.50 14.76 12.71 98.85 0.78 0.72 sand 70-80 42-44 25 UNDEFINED 4.75 15.58 24.17 176.25 0.73 0.76 sand >90 44-46 48 UNDEFINED 5.00 16.40 23.82 150.00 0.63 0.80 sand >90 44-46 48 UNDEFINED 5.25 17.22 21.54 182.25 0.85 0.85 sand >90 44-46 43 UNDEFINED 5.50 18.04 14.43 79.00 0.55 0.89 sand 70-80 42-44 29 UNDEFINED 5.75 18.86 9.53 53.95 0.57 0.92 sand to silty sand 60-70 40-42 24 UNDEFINED 6.00 19.69 9.69 68.00 0.70 0.93 sand to silty sand 60-70 40-42 24 UNDEFINED 6.25 20.51 18.35 98.00 0.53 0.95 sand 80-90 42-44 37 UNDEFINED 6.50 21.33 13.72 98.65 0.72 0.97 sand 70-80 42-44 27 UNDEFINED 6.75 22.15 7.83 55.05 0.70 0.98 sand to silty sand 60-70 40-42 20 UNDEFINED 7.00 22.97 9.06 43.80 0.48 1.00 sand to silty sand 60-70 40-42 23 UNDEFINED 7.25 23.79 11.94 49.75 0.42 1.02 sand 70-80 40-42 24 UNDEFINED 7.50 24.61 13.81 86.40 0.63 1.03 sand 70-80 42-44 28 UNDEFINED 7.75 25.43 13.44 73.44 0.55 1.05 sand 70-80 42-44 27 UNDEFINED 8.00 26.25 15.06 77.81 0.52 1.07 sand 70-80 42-44 30 UNDEFINED
347
Table C-2. continued----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 8.25 27.07 21.81 149.98 0.69 1.08 sand 80-90 44-46 44 UNDEFINED 8.50 27.89 20.63 113.80 0.55 1.10 sand 80-90 42-44 41 UNDEFINED 8.75 28.71 19.94 130.80 0.66 1.12 sand 80-90 42-44 40 UNDEFINED 9.00 29.53 13.75 95.50 0.69 1.13 sand 70-80 40-42 28 UNDEFINED 9.25 30.35 7.00 53.73 0.77 1.15 sand to silty sand 50-60 38-40 18 UNDEFINED 9.50 31.17 4.82 18.57 0.39 1.17 sand to silty sand 40-50 36-38 12 UNDEFINED 9.75 31.99 13.43 26.18 0.19 1.18 sand 70-80 40-42 27 UNDEFINED 10.00 32.81 14.59 37.75 0.26 1.20 sand 70-80 40-42 29 UNDEFINED 10.25 33.63 19.11 62.00 0.32 1.22 sand 80-90 42-44 38 UNDEFINED 10.50 34.45 21.75 138.75 0.64 1.23 sand 80-90 42-44 44 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 10
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
Figure C-2. Sheet pile section Larssen 4/20 at Hatfield site.
Figure C-3. Locations of inclinometer casings I0, I1, I5 at Hatfield site.
349
Table C-3. Soil properties from sounding CPT2University of Florida
Operator : A. Kort CPT Date : Before May 15, 1998 On Site Loc: Rotterdam Cone Used :156 Job No. : CPT02 Water table (meters) : 2 Tot. Unit Wt. (avg) : 15.5 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 0.00 0.00 0.00 0.02 undefined UNDFND UNDFD UDF UNDEFINED 0.50 1.64 0.00 0.00 0.00 0.06 undefined UNDFND UNDFD UDF UNDEFINED 0.75 2.46 0.00 0.00 0.00 0.10 undefined UNDFND UNDFD UDF UNDEFINED 1.00 3.28 0.00 0.00 0.00 0.14 undefined UNDFND UNDFD UDF UNDEFINED 1.25 4.10 0.00 0.00 0.00 0.17 undefined UNDFND UNDFD UDF UNDEFINED 1.50 4.92 0.00 0.00 0.00 0.21 undefined UNDFND UNDFD UDF UNDEFINED 1.75 5.74 0.48 13.65 2.87 0.25 clay UNDFND UNDFD 5 .4 2.00 6.56 1.53 43.24 2.83 0.29 clayey silt to silty clay UNDFND UNDFD 8 1.4 2.25 7.38 0.80 43.70 5.45 0.32 clay UNDFND UNDFD 8 .7 2.50 8.20 0.84 36.86 4.41 0.33 clay UNDFND UNDFD 8 .7 2.75 9.02 0.88 55.66 6.34 0.35 clay UNDFND UNDFD 9 .8 3.00 9.84 0.75 27.95 3.72 0.36 clay UNDFND UNDFD 8 .7 3.25 10.66 0.63 19.53 3.11 0.37 clay UNDFND UNDFD 6 .5 3.50 11.48 0.51 19.12 3.75 0.39 clay UNDFND UNDFD 5 .4 3.75 12.30 0.44 20.06 4.58 0.40 clay UNDFND UNDFD 4 .3 4.00 13.12 0.38 19.27 5.12 0.42 clay UNDFND UNDFD 4 .3 4.25 13.94 0.31 15.55 4.95 0.43 clay UNDFND UNDFD 3 .2 4.50 14.76 0.30 15.26 5.16 0.45 clay UNDFND UNDFD 3 .2 4.75 15.58 1.00 23.97 2.40 0.46 silty clay to clay UNDFND UNDFD 7 .9 5.00 16.40 0.50 18.25 3.65 0.47 clay UNDFND UNDFD 5 .4 5.25 17.22 1.09 22.08 2.03 0.49 clayey silt to silty clay UNDFND UNDFD 5 1.0 5.50 18.04 0.61 52.72 8.64 0.50 undefined UNDFND UNDFD UDF UNDEFINED 5.75 18.86 0.44 47.98 10.85 0.52 undefined UNDFND UNDFD UDF UNDEFINED 6.00 19.69 0.48 49.60 10.25 0.53 undefined UNDFND UNDFD UDF UNDEFINED 6.25 20.51 0.49 51.14 10.39 0.54 undefined UNDFND UNDFD UDF UNDEFINED 6.50 21.33 0.48 57.31 11.99 0.56 undefined UNDFND UNDFD UDF UNDEFINED 6.75 22.15 0.46 51.29 11.05 0.57 undefined UNDFND UNDFD UDF UNDEFINED 7.00 22.97 0.45 49.96 11.10 0.59 undefined UNDFND UNDFD UDF UNDEFINED 7.25 23.79 0.44 49.03 11.24 0.60 undefined UNDFND UNDFD UDF UNDEFINED 7.50 24.61 0.42 45.61 10.76 0.62 undefined UNDFND UNDFD UDF UNDEFINED 7.75 25.43 0.41 34.12 8.36 0.63 undefined UNDFND UNDFD UDF UNDEFINED 8.00 26.25 0.45 37.89 8.38 0.64 undefined UNDFND UNDFD UDF UNDEFINED
350
Table C-3. continued----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 8.25 27.07 0.43 38.74 8.97 0.66 undefined UNDFND UNDFD UDF UNDEFINED 8.50 27.89 0.40 28.07 7.02 0.67 clay UNDFND UNDFD 4 .2 8.75 28.71 0.40 35.12 8.78 0.69 undefined UNDFND UNDFD UDF UNDEFINED 9.00 29.53 0.40 32.50 8.13 0.70 undefined UNDFND UNDFD UDF UNDEFINED 9.25 30.35 0.40 29.30 7.32 0.72 clay UNDFND UNDFD 4 .2 9.50 31.17 0.40 27.95 6.99 0.73 clay UNDFND UNDFD 4 .2 9.75 31.99 0.40 27.25 6.81 0.74 clay UNDFND UNDFD 4 .2 10.00 32.81 0.40 29.60 7.40 0.76 clay UNDFND UNDFD 4 .2 10.25 33.63 0.40 28.70 7.17 0.77 clay UNDFND UNDFD 4 .2 10.50 34.45 0.40 34.40 8.60 0.79 undefined UNDFND UNDFD UDF UNDEFINED 10.75 35.27 0.40 20.78 5.19 0.80 clay UNDFND UNDFD 4 .2 11.00 36.09 0.40 17.40 4.35 0.81 clay UNDFND UNDFD 4 .2 11.25 36.91 0.40 15.59 3.90 0.83 clay UNDFND UNDFD 4 .2 11.50 37.73 0.40 12.67 3.17 0.84 clay UNDFND UNDFD 4 .2 11.75 38.55 0.40 12.88 3.22 0.86 clay UNDFND UNDFD 4 .2 12.00 39.37 0.40 14.43 3.61 0.87 clay UNDFND UNDFD 4 .2 12.25 40.19 0.40 13.45 3.36 0.89 clay UNDFND UNDFD 4 .2 12.50 41.01 0.40 9.83 2.46 0.90 clay UNDFND UNDFD 4 .2 12.75 41.83 0.40 12.13 3.03 0.91 clay UNDFND UNDFD 4 .2 13.00 42.65 0.40 12.00 3.00 0.93 clay UNDFND UNDFD 4 .2 13.25 43.47 0.55 11.52 2.09 0.94 silty clay to clay UNDFND UNDFD 4 .3 13.50 44.29 0.50 12.15 2.45 0.96 clay UNDFND UNDFD 5 .2 13.75 45.11 0.49 17.73 3.63 0.97 clay UNDFND UNDFD 5 .2 14.00 45.93 0.48 17.10 3.58 0.99 clay UNDFND UNDFD 5 .2 14.25 46.75 0.47 16.11 3.44 1.00 clay UNDFND UNDFD 5 .2 14.50 47.57 0.46 15.13 3.29 1.01 clay UNDFND UNDFD 5 .2 14.75 48.39 0.45 14.19 3.15 1.03 clay UNDFND UNDFD 5 .2 15.00 49.21 0.44 13.27 3.02 1.04 clay UNDFND UNDFD 4 .2 15.25 50.03 0.43 12.38 2.87 1.06 clay UNDFND UNDFD 4 .1 15.50 50.85 0.42 11.52 2.73 1.07 clay UNDFND UNDFD 4 .1 15.75 51.67 0.41 10.68 2.59 1.09 clay UNDFND UNDFD 4 .1 16.00 52.49 0.40 11.77 2.91 1.10 clay UNDFND UNDFD 4 .1 16.25 53.31 0.46 18.29 3.98 1.11 clay UNDFND UNDFD 5 .2 16.50 54.13 0.56 26.45 4.72 1.13 clay UNDFND UNDFD 6 .3 16.75 54.95 0.55 20.71 3.77 1.14 clay UNDFND UNDFD 6 .2 17.00 55.77 0.63 20.47 3.27 1.16 clay UNDFND UNDFD 6 .3 17.25 56.59 0.58 32.59 5.66 1.17 clay UNDFND UNDFD 6 .3 17.50 57.41 0.88 9.08 1.03 1.18 clayey silt to silty clay UNDFND UNDFD 4 .6 17.75 58.23 3.69 26.07 0.71 1.20 silty sand to sandy silt <40 34-36 12 UNDEFINED 18.00 59.06 5.30 42.34 0.80 1.21 silty sand to sandy silt 40-50 36-38 18 UNDEFINED
351
Table C-3. continued----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 18.25 59.88 5.33 47.94 0.90 1.23 silty sand to sandy silt 40-50 36-38 18 UNDEFINED 18.50 60.70 5.15 63.99 1.24 1.24 silty sand to sandy silt 40-50 36-38 17 UNDEFINED 18.75 61.52 5.41 53.71 0.99 1.26 silty sand to sandy silt 40-50 36-38 18 UNDEFINED 19.00 62.34 8.80 56.41 0.64 1.27 sand to silty sand 60-70 38-40 22 UNDEFINED 19.25 63.16 7.68 62.88 0.82 1.28 sand to silty sand 50-60 38-40 19 UNDEFINED 19.50 63.98 4.96 77.26 1.56 1.30 silty sand to sandy silt 40-50 36-38 17 UNDEFINED 19.75 64.80 2.94 26.02 0.89 1.31 silty sand to sandy silt <40 32-34 10 UNDEFINED 20.00 65.62 5.95 62.99 1.06 1.33 silty sand to sandy silt 40-50 36-38 20 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 10
352
Figure C-4. Pile sections AZ13 and L607K with casings at Rotterdam site.
353
APPENDIX DGREEN COVE SPRINGS SHALLOW FOOTING
354
Table D-1. Soil Properties from CPT sounding, Green Cove Springs----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 12.03 56.08 0.47 0.02 sand >90 >48 24 UNDEFINED 0.50 1.64 10.12 63.04 0.62 0.07 sand to silty sand >90 >48 25 UNDEFINED 0.75 2.46 13.96 77.20 0.55 0.11 sand >90 >48 28 UNDEFINED 1.00 3.28 9.44 65.64 0.70 0.16 sand to silty sand >90 >48 24 UNDEFINED 1.25 4.10 11.06 74.24 0.67 0.20 sand to silty sand >90 >48 28 UNDEFINED 1.50 4.92 11.59 44.48 0.38 0.25 sand >90 46-48 23 UNDEFINED 1.75 5.74 9.03 43.46 0.48 0.29 sand to silty sand 80-90 44-46 23 UNDEFINED 2.00 6.56 8.27 57.74 0.70 0.34 sand to silty sand 70-80 44-46 21 UNDEFINED 2.25 7.38 5.12 38.10 0.74 0.38 silty sand to sandy silt 60-70 42-44 17 UNDEFINED 2.50 8.20 1.52 10.38 0.68 0.43 sandy silt to clayey silt UNDFND UNDFD 6 .9 2.75 9.02 3.36 17.58 0.52 0.47 silty sand to sandy silt 40-50 38-40 11 UNDEFINED 3.00 9.84 6.40 27.98 0.44 0.52 sand to silty sand 60-70 40-42 16 UNDEFINED 3.25 10.66 5.14 33.98 0.66 0.56 sand to silty sand 50-60 40-42 13 UNDEFINED 3.50 11.48 5.50 32.76 0.60 0.61 sand to silty sand 50-60 40-42 14 UNDEFINED 3.75 12.30 10.51 57.42 0.55 0.65 sand 70-80 42-44 21 UNDEFINED 4.00 13.12 10.58 68.26 0.65 0.70 sand to silty sand 70-80 42-44 26 UNDEFINED 4.25 13.94 6.73 44.74 0.66 0.74 sand to silty sand 60-70 40-42 17 UNDEFINED 4.50 14.76 3.16 21.16 0.67 0.79 silty sand to sandy silt <40 36-38 11 UNDEFINED 4.75 15.58 5.08 22.60 0.44 0.83 sand to silty sand 50-60 38-40 13 UNDEFINED 5.00 16.40 3.67 14.24 0.39 0.88 silty sand to sandy silt 40-50 36-38 12 UNDEFINED 5.25 17.22 3.10 3.76 0.12 0.92 silty sand to sandy silt <40 34-36 10 UNDEFINED 5.50 18.04 1.67 6.04 0.36 0.97 sandy silt to clayey silt UNDFND UNDFD 7 1.0 5.75 18.86 1.08 4.40 0.41 1.01 sandy silt to clayey silt UNDFND UNDFD 4 .6 6.00 19.69 0.48 5.04 1.06 1.06 sensitive fine grained UNDFND UNDFD 2 .2 6.25 20.51 0.45 3.66 0.81 1.10 sensitive fine grained UNDFND UNDFD 2 .2 6.50 21.33 0.50 5.68 1.14 1.15 sensitive fine grained UNDFND UNDFD 2 .2 6.75 22.15 0.46 4.68 1.02 1.19 sensitive fine grained UNDFND UNDFD 2 .2 7.00 22.97 0.52 3.94 0.76 1.24 sensitive fine grained UNDFND UNDFD 3 .2 7.25 23.79 0.64 7.20 1.13 1.28 sensitive fine grained UNDFND UNDFD 3 .3 7.50 24.61 0.68 7.94 1.17 1.33 sensitive fine grained UNDFND UNDFD 3 .3 7.75 25.43 0.68 8.86 1.30 1.37 undefined UNDFND UNDFD UDF UNDEFINED 8.00 26.25 0.71 8.58 1.21 1.42 undefined UNDFND UNDFD UDF UNDEFINED 8.25 27.07 0.76 8.74 1.14 1.46 clayey silt to silty clay UNDFND UNDFD 4 .4 8.50 27.89 0.89 10.18 1.14 1.51 clayey silt to silty clay UNDFND UNDFD 4 .4 8.75 28.71 0.85 6.84 0.81 1.55 clayey silt to silty clay UNDFND UNDFD 4 .4 9.00 29.53 0.83 7.60 0.92 1.60 clayey silt to silty clay UNDFND UNDFD 4 .4 9.25 30.35 0.82 9.20 1.12 1.64 clayey silt to silty clay UNDFND UNDFD 4 .4 9.50 31.17 0.88 8.98 1.02 1.69 clayey silt to silty clay UNDFND UNDFD 4 .4 9.75 31.99 0.84 8.36 0.99 1.73 clayey silt to silty clay UNDFND UNDFD 4 .4 10.00 32.81 0.97 4.26 0.44 1.78 sandy silt to clayey silt UNDFND UNDFD 4 .5 10.25 33.63 0.84 4.34 0.52 1.82 sensitive fine grained UNDFND UNDFD 4 .4 10.50 34.45 0.91 11.02 1.21 1.87 clayey silt to silty clay UNDFND UNDFD 5 .4----------------------------------------------------------------------------------------------------------------------------------- Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 15
355
Table D-2. Soil Properties from DMT sounding, Green Cove SpringsDILATOMETER DATA LISTING & INTERPRETATION (BASED ON THE 1988 DILATOMETER MANUAL) SNDG. NO. DMT-22GPE, INC.JOB FILE: UNIVERSITY OF FLORIDA RESEARCH FILE NO. : 83-500LOCATION: GREEN COVE SPRINGSSNDG.BY : BRIAN/CHRIS/LANDY - UF CPT TRUCK SNDG.DATE: 12 DEC 01ANAL.BY : LANDY R./UF - GEOTECHNICAL GROUP ANAL.DATE: 12 DEC 01
ANALYSIS PARAMETERS: LO RANGE =10.00 BARS ROD DIAM. = 3.70 CM BL.THICK. = 15.0 MM SU FACTOR = 1.00 SURF.ELEV. = 0.00 M LO GAGE 0 = 0.00 BARS FR.RED.DIA. = 4.80 CM BL.WIDTH = 96.0 MM PHI FACTOR = 1.00 WATER DEPTH = 1.68 M HI GAGE 0 = 0.00 BARS LIN.ROD WT. = 6.50 KGF/M DELTA-A = 0.20 BARS OCR FACTOR = 1.00 SP.GR.WATER = 1.000 CAL GAGE 0 = 0.00 BARS DELTA/PHI = 0.50 DELTA-B = 0.27 BARS M FACTOR = 1.00 MAX SU ID = 0.60 SU OPTION = MARCHETTI MIN PHI ID = 1.20 OCR OPTION= MARCHETTI K0 FACTOR = 1.00UNIT CONVERSIONS: 1 BAR = 1.019 KGF/CM2 = 100 KPA = 1.044 TSF = 14.51 PSI 1 M = 3.2808 FT
Z THRUST A B DA DB P0 P1 U0 GAMMA SVP KD ID ED K0 SU QD PHI SIGFF PHIO PC OCR M SOIL TYPE (M) (KGF) (BAR) (BAR) (BAR) (BAR) (BAR) (BAR) (BAR) (T/M3) (BAR) (BAR) (BAR) (BAR) (DEG) (BAR) (DEG) (BAR) (BAR)***** ****** ***** ***** ***** ***** ***** ***** ****** ****** ****** ***** ***** ****** ***** ***** ***** ***** ****** ***** ***** ***** ****** ************
0.20 463. 0.15 15.00 0.20 0.27 0.35X 14.73 0.000 1.80 0.038 9.21 41.09 499. 1212. SAND 0.40 1399. 0.45 5.95 0.20 0.27 0.40 5.68 0.000 1.70 0.072 5.51 13.25 183. 363. SAND 0.60 1919. 2.25 15.60 0.20 0.27 1.81 15.33 0.000 1.90 0.108 16.77 7.49 469. 1403. SAND 0.80 2459. 4.25 15.80 0.20 0.27 3.90 15.53 0.000 2.00 0.146 26.69 2.99 404. 3.10 73.4 45.2 0.25 41.9 9.67 66.2 1385. SILTY SAND 1.00 1713. 3.45 13.80 0.20 0.27 3.16 13.53 0.000 1.90 0.184 17.13 3.29 360. 2.07 49.7 43.1 0.31 39.9 5.59 30.3 1083. SILTY SAND 1.20 4136. 4.45 19.80 0.20 0.27 3.91 19.53 0.000 2.00 0.222 17.56 4.00 542. 1.80 137.1 47.5 0.39 45.1 5.31 23.9 1644. SAND 1.40 6322. 6.55 24.40 0.20 0.27 5.88 24.13 0.000 2.00 0.262 22.47 3.10 633. 2.37 209.6 47.7 0.46 45.5 10.39 39.7 2068. SILTY SAND 1.60 6337. 6.65 26.80 0.20 0.27 5.87 26.53 0.000 2.00 0.301 19.49 3.52 717. 2.04 210.6 47.5 0.52 45.4 9.04 30.0 2246. SAND 1.80 6466. 8.25 17.40 0.20 0.27 8.02 17.13 0.012 1.95 0.328 24.40 1.14 316. 3.11 16.24 49.5 1058. SILT 2.00 5211. 5.85 19.40 0.20 0.27 5.40 19.13 0.031 2.00 0.347 15.45 2.56 477. 1.66 170.9 46.3 0.60 44.4 7.05 20.3 1388. SILTY SAND 2.20 4547. 4.65 18.00 0.20 0.27 4.21 17.73 0.051 2.00 0.367 11.33 3.25 469. 1.15 152.4 46.5 0.63 44.6 3.74 10.2 1229. SILTY SAND 2.40 3343. 3.65 13.20 0.20 0.27 3.40 12.93 0.071 1.90 0.385 8.63 2.87 331. 0.92 111.4 45.2 0.66 43.3 2.51 6.5 784. SILTY SAND 2.60 1337. 2.60 8.85 0.20 0.27 2.51 8.58 0.090 1.90 0.403 6.01 2.51 211. 0.86 40.7 39.6 0.66 37.4 1.99 4.9 430. SILTY SAND 2.80 1538. 2.20 8.70 0.20 0.27 2.10 8.43 0.110 1.90 0.421 4.73 3.18 220. 0.65 50.3 41.2 0.70 39.1 1.24 2.9 406. SILTY SAND 3.00 1250. 1.95 8.15 0.20 0.27 1.86 7.88 0.130 1.80 0.437 3.96 3.47 209. 0.61 40.9 39.8 0.72 37.7 1.06 2.4 354. SAND 3.20 1132. 1.70 7.15 0.20 0.27 1.65 6.88 0.149 1.80 0.453 3.31 3.48 181. 0.55 37.6 39.3 0.74 37.2 0.87 1.9 280. SAND 3.40 1116. 1.60 6.80 0.20 0.27 1.56 6.53 0.169 1.80 0.469 2.97 3.56 172. 0.51 37.6 39.2 0.76 37.1 0.77 1.6 249. SAND 3.60 1996. 2.05 8.30 0.20 0.27 1.96 8.03 0.188 1.80 0.485 3.66 3.42 211. 0.46 69.0 42.7 0.81 41.0 0.75 1.5 343. SAND 3.80 3014. 3.65 13.20 0.20 0.27 3.40 12.93 0.208 1.90 0.501 6.36 2.99 331. 0.74 100.4 43.7 0.85 42.1 2.03 4.0 697. SILTY SAND 4.00 2752. 3.95 13.40 0.20 0.27 3.70 13.13 0.228 1.90 0.519 6.69 2.71 327. 0.83 89.1 42.7 0.87 41.0 2.57 5.0 701. SILTY SAND 4.20 1296. 2.30 7.90 0.20 0.27 2.24 7.63 0.247 1.80 0.536 3.73 2.70 187. 0.61 41.9 38.7 0.87 36.8 1.28 2.4 303. SILTY SAND 4.40 710. 1.30 5.75 0.20 0.27 1.30 5.48 0.267 1.80 0.551 1.88 4.04 145. 0.48 24.4 35.4 0.87 33.3 0.67 1.2 152. SAND 4.60 1235. 2.10 8.55 0.20 0.27 2.00 8.28 0.287 1.80 0.567 3.02 3.66 218. 0.54 41.1 38.4 0.92 36.6 1.01 1.8 318. SAND 4.80 1029. 1.80 7.35 0.20 0.27 1.75 7.08 0.306 1.80 0.583 2.47 3.70 185. 0.50 34.7 37.3 0.94 35.5 0.86 1.5 238. SAND 5.00 890. 1.55 6.75 0.20 0.27 1.51 6.48 0.326 1.80 0.598 1.98 4.18 172. 0.47 30.7 36.4 0.95 34.6 0.72 1.2 189. SAND 5.20 993. 1.80 8.25 0.20 0.27 1.70 7.98 0.345 1.80 0.614 2.21 4.63 218. 0.48 33.8 36.8 0.98 35.1 0.81 1.3 259. SAND 5.40 1003. 1.70 5.70 0.20 0.27 1.72 5.43 0.365 1.80 0.630 2.16 2.73 129. 0.48 34.3 36.8 1.01 35.0 0.82 1.3 147. SILTY SAND 5.60 432. 1.75 3.80 0.20 0.27 1.87 3.53 0.385 1.60 0.643 2.31 1.12 58. 0.62 0.81 1.3 60. SILT 5.80 339. 2.20 3.15 0.20 0.27 2.38 2.88 0.404 1.60 0.655 3.01 0.26 17. 0.79 0.24 1.24 1.9 22. CLAY 6.00 231. 2.40 3.35 0.20 0.27 2.58 3.08 0.424 1.60 0.667 3.23 0.23 17. 0.83 0.27 1.41 2.1 23. CLAY 6.20 247. 2.60 3.90 0.20 0.27 2.76 3.63 0.444 1.60 0.679 3.41 0.38 30. 0.87 0.29 1.56 2.3 42. SILTY CLAY 6.40 273. 2.80 4.15 0.20 0.27 2.96 3.88 0.463 1.70 0.692 3.60 0.37 32. 0.91 0.32 1.73 2.5 47. SILTY CLAY 6.80 329. 2.70 4.05 0.20 0.27 2.86 3.78 0.502 1.70 0.719 3.27 0.39 32. 0.84 0.29 1.55 2.2 43. SILTY CLAY 7.20 370. 3.72 5.20 0.20 0.27 3.87 4.93 0.542 1.70 0.747 4.46 0.32 37. 1.07 0.45 2.61 3.5 62. CLAY 7.60 417. 3.50 5.25 0.20 0.27 3.64 4.98 0.581 1.70 0.774 3.95 0.44 47. 0.98 0.40 2.24 2.9 72. SILTY CLAY 8.00 381. 3.85 5.45 0.20 0.27 3.99 5.18 0.620 1.70 0.801 4.21 0.35 41. 1.02 0.45 2.56 3.2 66. SILTY CLAY 8.40 406. 4.25 5.70 0.20 0.27 4.40 5.43 0.659 1.70 0.829 4.51 0.28 36. 1.08 0.50 2.95 3.6 60. CLAY****************************************************************************************************************************************************************************
356
PMT at depth 2.0m
0.002.004.006.008.00
10.0012.0014.0016.00
-20 0 20 40 60 80 100
Volume (cm3)
Pres
sure
(bar
) Insitu
Plaxis HS
PMT at depth 3.0m
0.00
2.00
4.00
6.00
8.00
10.00
12.00
14.00
16.00
-20 0 20 40 60 80 100Volume (cm3)
Pres
sure
(bar
) Insitu
Plaxis HS
Figure D-1. PMT curve-fitting for Green Cove Springs site
357
PMT at depth 5.0m
0.00
2.00
4.00
6.00
8.00
10.00
12.00
14.00
16.00
-20 0 20 40 60 80 100Volume (cm3)
Pres
sure
(bar
) Insitu
Plaxis HS
PMT at depth 7.0m
0.00
2.00
4.00
6.00
8.00
10.00
12.00
14.00
16.00
-20 0 20 40 60 80Volume (cm3)
Pres
sure
(bar
) Insitu
Plaxis HS
Figure D-1. continued
358
PMT at depth 10.0m
0.00
2.00
4.00
6.00
8.00
10.00
12.00
14.00
16.00
-20 0 20 40 60 80 100Volume (cm3)
Pres
sure
(bar
) Insitu
Plaxis HS
Figure D-1. continued.
359
APPENDIX ETEXAS A&M FOOTINGS AND UTAH MAT FOUNDATION
360
Table E-1. Soil Properties from CPT2 data reduction, Texas A&M University
University of Florida
Operator :JLB/PJ CPT Date :11-01-02 16:00 On Site Loc:Texas A&M Univ. Cone Used :156 Job No. :CPT02 Water table (meters) : 4.9 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 1.20 7.34 0.61 0.02 sandy silt to clayey silt UNDFND UNDFD 5 .7 0.50 1.64 2.71 22.35 0.82 0.07 silty sand to sandy silt 60-70 46-48 9 UNDEFINED 0.75 2.46 3.46 22.03 0.64 0.11 silty sand to sandy silt 60-70 44-46 12 UNDEFINED 1.00 3.28 4.07 28.74 0.71 0.16 silty sand to sandy silt 60-70 44-46 14 UNDEFINED 1.25 4.10 4.84 38.79 0.80 0.20 silty sand to sandy silt 60-70 44-46 16 UNDEFINED 1.50 4.92 8.33 80.46 0.97 0.25 sand to silty sand 80-90 46-48 21 UNDEFINED 1.75 5.74 9.15 76.82 0.84 0.29 sand to silty sand 80-90 44-46 23 UNDEFINED 2.00 6.56 5.99 28.74 0.48 0.34 sand to silty sand 60-70 42-44 15 UNDEFINED 2.25 7.38 8.74 40.99 0.47 0.38 sand to silty sand 70-80 44-46 22 UNDEFINED 2.50 8.20 7.21 55.46 0.77 0.43 sand to silty sand 70-80 42-44 18 UNDEFINED 2.75 9.02 5.67 33.43 0.59 0.47 sand to silty sand 60-70 40-42 14 UNDEFINED 3.00 9.84 7.57 45.98 0.61 0.52 sand to silty sand 60-70 42-44 19 UNDEFINED 3.25 10.66 10.37 66.71 0.64 0.56 sand to silty sand 70-80 42-44 26 UNDEFINED 3.50 11.48 11.78 72.61 0.62 0.61 sand 70-80 42-44 24 UNDEFINED 3.75 12.30 12.70 74.07 0.58 0.65 sand 80-90 42-44 25 UNDEFINED 4.00 13.12 12.93 84.77 0.66 0.70 sand 70-80 42-44 26 UNDEFINED 4.25 13.94 12.35 90.92 0.74 0.74 sand 70-80 42-44 25 UNDEFINED 4.50 14.76 10.76 74.23 0.69 0.79 sand to silty sand 70-80 42-44 27 UNDEFINED 4.75 15.58 11.72 87.36 0.75 0.83 sand to silty sand 70-80 42-44 29 UNDEFINED 5.00 16.40 10.14 115.90 1.14 0.88 sand to silty sand 60-70 40-42 25 UNDEFINED 5.25 17.22 8.05 144.64 1.80 0.90 silty sand to sandy silt 60-70 40-42 27 UNDEFINED 5.50 18.04 9.72 131.23 1.35 0.92 sand to silty sand 60-70 40-42 24 UNDEFINED 5.75 18.86 6.59 159.77 2.42 0.94 sandy silt to clayey silt UNDFND UNDFD 26 4.3 6.00 19.69 4.60 138.89 3.02 0.96 sandy silt to clayey silt UNDFND UNDFD 18 2.9 6.25 20.51 8.38 177.20 2.11 0.98 silty sand to sandy silt 60-70 40-42 28 UNDEFINED 6.50 21.33 8.35 192.34 2.30 1.00 silty sand to sandy silt 60-70 40-42 28 UNDEFINED 6.75 22.15 8.02 89.04 1.11 1.02 sand to silty sand 60-70 38-40 20 UNDEFINED 7.00 22.97 7.61 75.86 1.00 1.04 sand to silty sand 50-60 38-40 19 UNDEFINED 7.25 23.79 7.15 72.26 1.01 1.06 sand to silty sand 50-60 38-40 18 UNDEFINED 7.50 24.61 6.06 54.28 0.90 1.08 sand to silty sand 50-60 38-40 15 UNDEFINED 7.75 25.43 7.60 67.18 0.88 1.11 sand to silty sand 50-60 38-40 19 UNDEFINED
361
Table E-1. continued
----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 8.00 26.25 3.55 77.20 2.18 1.13 sandy silt to clayey silt UNDFND UNDFD 14 2.2 8.25 27.07 2.67 86.78 3.25 1.15 clayey silt to silty clay UNDFND UNDFD 13 1.6 8.50 27.89 5.56 91.00 1.64 1.17 silty sand to sandy silt 40-50 36-38 19 UNDEFINED 8.75 28.71 4.18 111.11 2.66 1.19 sandy silt to clayey silt UNDFND UNDFD 17 2.6 9.00 29.53 5.35 250.96 4.69 1.21 silty clay to clay UNDFND UNDFD 36 3.4 9.25 30.35 6.73 231.80 3.44 1.23 sandy silt to clayey silt UNDFND UNDFD 27 4.3 9.50 31.17 6.78 299.33 4.41 1.25 clayey silt to silty clay UNDFND UNDFD 34 4.4 9.75 31.99 8.85 262.93 2.97 1.27 sandy silt to clayey silt UNDFND UNDFD 35 5.7 10.00 32.81 19.30 234.67 1.22 1.29 sand 80-90 42-44 39 UNDEFINED 10.25 33.63 11.20 303.54 2.71 1.31 sandy silt to clayey silt UNDFND UNDFD 45 7.3 10.50 34.45 9.51 343.73 3.61 1.33 sandy silt to clayey silt UNDFND UNDFD 38 6.2 10.75 35.27 9.58 208.00 2.17 1.35 silty sand to sandy silt 60-70 38-40 32 UNDEFINED 11.00 36.09 8.52 172.89 2.03 1.37 silty sand to sandy silt 50-60 38-40 28 UNDEFINED 11.25 36.91 8.95 226.05 2.53 1.39 sandy silt to clayey silt UNDFND UNDFD 36 5.8 11.50 37.73 9.00 210.73 2.34 1.41 silty sand to sandy silt 50-60 38-40 30 UNDEFINED 11.75 38.55 8.88 206.90 2.33 1.43 silty sand to sandy silt 50-60 38-40 30 UNDEFINED 12.00 39.37 8.76 214.56 2.45 1.45 silty sand to sandy silt 50-60 38-40 29 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 15
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
362
Table E-2. Soil Properties from CPT5 data reduction, Texas A&M University
University of Florida
Operator :JLB/PJ CPT Date :11-01-02 20:45 On Site Loc:Texas A&M Univ. Cone Used :156 Job No. :CPT05 Water table (meters) : 4.9 Tot. Unit Wt. (avg) : 18 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 1.44 8.62 0.60 0.02 sandy silt to clayey silt UNDFND UNDFD 6 .9 0.50 1.64 3.26 18.20 0.56 0.07 silty sand to sandy silt 70-80 46-48 11 UNDEFINED 0.75 2.46 7.18 52.87 0.74 0.11 sand to silty sand 80-90 >48 18 UNDEFINED 1.00 3.28 9.88 51.01 0.52 0.16 sand to silty sand >90 >48 25 UNDEFINED 1.25 4.10 9.33 46.50 0.50 0.20 sand to silty sand 80-90 46-48 23 UNDEFINED 1.50 4.92 12.17 79.70 0.66 0.25 sand >90 46-48 24 UNDEFINED 1.75 5.74 10.86 70.88 0.65 0.29 sand to silty sand 80-90 46-48 27 UNDEFINED 2.00 6.56 9.55 51.25 0.54 0.34 sand to silty sand 80-90 44-46 24 UNDEFINED 2.25 7.38 8.53 44.35 0.52 0.38 sand to silty sand 70-80 44-46 21 UNDEFINED 2.50 8.20 9.16 51.30 0.56 0.43 sand to silty sand 70-80 44-46 23 UNDEFINED 2.75 9.02 7.06 40.66 0.58 0.47 sand to silty sand 60-70 42-44 18 UNDEFINED 3.00 9.84 8.18 50.36 0.62 0.52 sand to silty sand 70-80 42-44 20 UNDEFINED 3.25 10.66 8.90 64.59 0.73 0.56 sand to silty sand 70-80 42-44 22 UNDEFINED 3.50 11.48 9.34 59.87 0.64 0.61 sand to silty sand 70-80 42-44 23 UNDEFINED 3.75 12.30 9.23 52.68 0.57 0.65 sand to silty sand 70-80 42-44 23 UNDEFINED 4.00 13.12 8.89 52.87 0.59 0.70 sand to silty sand 60-70 42-44 22 UNDEFINED 4.25 13.94 9.57 69.54 0.73 0.74 sand to silty sand 70-80 42-44 24 UNDEFINED 4.50 14.76 9.18 79.69 0.87 0.79 sand to silty sand 60-70 40-42 23 UNDEFINED 4.75 15.58 9.84 61.30 0.62 0.83 sand to silty sand 60-70 40-42 25 UNDEFINED 5.00 16.40 11.68 78.93 0.68 0.88 sand 70-80 42-44 23 UNDEFINED 5.25 17.22 12.74 100.26 0.79 0.90 sand to silty sand 70-80 42-44 32 UNDEFINED 5.50 18.04 6.15 126.76 2.06 0.92 silty sand to sandy silt 50-60 38-40 21 UNDEFINED 5.75 18.86 8.62 95.78 1.11 0.94 sand to silty sand 60-70 40-42 22 UNDEFINED 6.00 19.69 8.24 169.54 2.06 0.96 silty sand to sandy silt 60-70 40-42 27 UNDEFINED 6.25 20.51 9.05 115.90 1.28 0.98 sand to silty sand 60-70 40-42 23 UNDEFINED 6.50 21.33 3.74 170.98 4.57 1.00 silty clay to clay UNDFND UNDFD 25 2.4 6.75 22.15 5.27 158.05 3.00 1.02 sandy silt to clayey silt UNDFND UNDFD 21 3.4 7.00 22.97 7.98 130.75 1.64 1.04 silty sand to sandy silt 60-70 38-40 27 UNDEFINED 7.25 23.79 10.18 102.49 1.01 1.06 sand to silty sand 60-70 40-42 25 UNDEFINED 7.50 24.61 6.39 111.59 1.75 1.08 silty sand to sandy silt 50-60 38-40 21 UNDEFINED 7.75 25.43 6.76 113.51 1.68 1.11 silty sand to sandy silt 50-60 38-40 23 UNDEFINED 8.00 26.25 4.63 141.28 3.05 1.13 sandy silt to clayey silt UNDFND UNDFD 19 2.9 8.25 27.07 5.88 111.11 1.89 1.15 silty sand to sandy silt 50-60 36-38 20 UNDEFINED 8.50 27.89 4.59 112.07 2.44 1.17 sandy silt to clayey silt UNDFND UNDFD 18 2.9
363
Table E-2. continued----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 8.75 28.71 2.68 108.56 4.05 1.19 silty clay to clay UNDFND UNDFD 18 1.6 9.00 29.53 5.97 50.45 0.84 1.21 sand to silty sand 40-50 36-38 15 UNDEFINED 9.25 30.35 5.41 144.63 2.67 1.23 sandy silt to clayey silt UNDFND UNDFD 22 3.4 9.50 31.17 9.16 215.84 2.36 1.25 silty sand to sandy silt 60-70 38-40 31 UNDEFINED 9.75 31.99 25.98 243.93 0.94 1.27 sand >90 44-46 >50 UNDEFINED 10.00 32.81 28.74 301.72 1.05 1.29 sand >90 44-46 >50 UNDEFINED 10.25 33.63 23.73 343.87 1.45 1.31 sand to silty sand 80-90 42-44 >50 UNDEFINED 10.50 34.45 9.03 268.20 2.97 1.33 sandy silt to clayey silt UNDFND UNDFD 36 5.8 10.75 35.27 9.38 236.40 2.52 1.35 silty sand to sandy silt 60-70 38-40 31 UNDEFINED 11.00 36.09 9.38 254.89 2.72 1.37 sandy silt to clayey silt UNDFND UNDFD 38 6.1 11.25 36.91 9.19 219.44 2.39 1.39 silty sand to sandy silt 60-70 38-40 31 UNDEFINED 11.50 37.73 9.00 237.55 2.64 1.41 sandy silt to clayey silt UNDFND UNDFD 36 5.8 11.75 38.55 8.80 226.06 2.57 1.43 sandy silt to clayey silt UNDFND UNDFD 35 5.7 12.00 39.37 8.61 233.24 2.71 1.45 sandy silt to clayey silt UNDFND UNDFD 34 5.5 12.25 40.19 8.41 262.45 3.12 1.47 sandy silt to clayey silt UNDFND UNDFD 34 5.4 12.50 41.01 8.22 250.96 3.05 1.49 sandy silt to clayey silt UNDFND UNDFD 33 5.3 12.75 41.83 8.42 229.89 2.73 1.51 sandy silt to clayey silt UNDFND UNDFD 34 5.4 13.00 42.65 8.10 217.12 2.68 1.54 sandy silt to clayey silt UNDFND UNDFD 32 5.2 13.25 43.47 8.24 239.30 2.91 1.56 sandy silt to clayey silt UNDFND UNDFD 33 5.3 13.50 44.29 8.37 282.09 3.37 1.58 sandy silt to clayey silt UNDFND UNDFD 33 5.4 13.75 45.11 8.51 216.99 2.55 1.60 sandy silt to clayey silt UNDFND UNDFD 34 5.5 14.00 45.93 16.28 328.22 2.02 1.62 silty sand to sandy silt 70-80 40-42 >50 UNDEFINED 14.25 46.75 16.74 314.18 1.88 1.64 sand to silty sand 70-80 40-42 42 UNDEFINED 14.50 47.57 9.50 181.99 1.92 1.66 silty sand to sandy silt 50-60 38-40 32 UNDEFINED 14.75 48.39 9.39 205.25 2.19 1.68 silty sand to sandy silt 50-60 38-40 31 UNDEFINED 15.00 49.21 9.07 191.57 2.11 1.70 silty sand to sandy silt 50-60 36-38 30 UNDEFINED 15.25 50.03 8.75 177.89 2.03 1.72 silty sand to sandy silt 50-60 36-38 29 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 15
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
364
Table E-3. Soil Properties from DMT3 data reduction, Texas A&M University
DILATOMETER DATA LISTING & INTERPRETATION (BASED ON THE 1988 DILATOMETER MANUAL) SNDG. NO. DMT-03GPE, INC.JOB FILE: UNIVERSITY OF FLORIDA RESEARCH FILE NO. : 93-5 FootingsLOCATION: Texas A&M UniversitySNDG.BY : Robert Gibbens - University of L'Aquilla SNDG.DATE: May 27, 1993ANAL.BY : Landy R./Univ. of Florida - Geotechnical Group ANAL.DATE: JAN 12, 2002 (Update
ANALYSIS PARAMETERS: LO RANGE =10.00 BARS ROD DIAM. = 3.70 CM BL.THICK. = 15.0 MM SU FACTOR = 1.00 SURF.ELEV. = 0.00 M LO GAGE 0 = 0.00 BARS FR.RED.DIA. = 4.80 CM BL.WIDTH = 96.0 MM PHI FACTOR = 1.00 WATER DEPTH = 4.90 M HI GAGE 0 = 0.00 BARS LIN.ROD WT. = 6.50 KGF/M DELTA-A = 0.00 BARS OCR FACTOR = 1.00 SP.GR.WATER = 1.000 CAL GAGE 0 = 0.00 BARS DELTA/PHI = 0.50 DELTA-B = 0.00 BARS M FACTOR = 1.00 MAX SU ID = 0.60 SU OPTION = MARCHETTI MIN PHI ID = 1.20 OCR OPTION= MARCHETTI K0 FACTOR = 1.00UNIT CONVERSIONS: 1 BAR = 1.019 KGF/CM2 = 100 KPA = 1.044 TSF = 14.51 PSI 1 M = 3.2808 FT
Z THRUST A B DA DB P0 P1 U0 GAMMA SVP KD ID ED K0 SU QD PHI SIGFF PHIO PC OCR M SOIL TYPE (M) (KGF) (BAR) (BAR) (BAR) (BAR) (BAR) (BAR) (BAR) (T/M3) (BAR) (BAR) (BAR) (BAR) (DEG) (BAR) (DEG) (BAR) (BAR)***** ****** ***** ***** ***** ***** ***** ***** ****** ****** ****** ***** ***** ****** ***** ***** ***** ***** ****** ***** ***** ***** ****** ************
1.00 2039. 3.35 10.40 0.00 0.00 3.00 10.40 0.000 1.90 0.180 16.65 2.47 257. 1.93 62.6 44.6 0.31 41.6 4.78 26.6 766. SILTY SAND 1.20 2039. 2.95 9.48 0.00 0.00 2.62 9.48 0.000 1.90 0.217 12.07 2.61 238. 1.37 64.8 44.7 0.37 41.9 3.00 13.8 637. SILTY SAND 1.40 1896. 2.39 7.75 0.00 0.00 2.12 7.75 0.000 1.90 0.255 8.34 2.65 195. 0.93 62.2 44.5 0.43 42.0 1.65 6.5 457. SILTY SAND 1.60 1967. 2.85 9.40 0.00 0.00 2.52 9.40 0.000 1.90 0.292 8.64 2.73 239. 1.02 63.1 43.6 0.49 41.2 2.21 7.6 566. SILTY SAND 1.80 2039. 2.95 9.90 0.00 0.00 2.60 9.90 0.000 1.90 0.329 7.91 2.80 253. 0.94 65.5 43.3 0.55 41.0 2.14 6.5 581. SILTY SAND 2.00 2212. 2.90 10.10 0.00 0.00 2.54 10.10 0.000 1.90 0.366 6.93 2.98 262. 0.81 72.6 43.6 0.62 41.5 1.79 4.9 572. SILTY SAND 2.20 2375. 4.10 12.51 0.00 0.00 3.68 12.51 0.000 1.90 0.404 9.11 2.40 306. 1.13 73.2 42.3 0.68 40.2 3.71 9.2 740. SILTY SAND 2.40 2406. 3.65 12.10 0.00 0.00 3.23 12.10 0.000 1.90 0.441 7.32 2.75 308. 0.90 76.8 42.6 0.74 40.7 2.60 5.9 685. SILTY SAND 2.60 2548. 4.53 14.29 0.00 0.00 4.04 14.29 0.000 1.90 0.478 8.45 2.54 356. 1.07 78.3 41.9 0.80 40.0 3.90 8.2 835. SILTY SAND 2.80 2599. 5.17 16.50 0.00 0.00 4.60 16.50 0.000 2.00 0.517 8.91 2.58 413. 1.15 77.8 41.2 0.86 39.5 4.83 9.3 989. SILTY SAND 3.00 2793. 5.15 15.70 0.00 0.00 4.62 15.70 0.000 2.00 0.556 8.32 2.40 384. 1.07 85.1 41.5 0.92 39.8 4.50 8.1 897. SILTY SAND 3.20 2752. 4.45 14.50 0.00 0.00 3.95 14.50 0.000 1.90 0.594 6.64 2.67 366. 0.86 87.0 41.7 0.99 40.2 3.11 5.2 782. SILTY SAND 3.40 2569. 5.00 15.89 0.00 0.00 4.46 15.89 0.000 2.00 0.632 7.05 2.57 397. 0.95 77.9 40.5 1.04 39.0 3.97 6.3 867. SILTY SAND 3.60 2559. 5.55 16.50 0.00 0.00 5.00 16.50 0.000 2.00 0.672 7.45 2.30 399. 1.03 75.3 39.8 1.10 38.4 4.83 7.2 890. SILTY SAND 3.80 2589. 5.49 16.53 0.00 0.00 4.94 16.53 0.000 2.00 0.711 6.95 2.35 402. 0.97 76.9 39.7 1.17 38.4 4.53 6.4 872. SILTY SAND 4.00 2854. 6.55 20.50 0.00 0.00 5.85 20.50 0.000 2.00 0.750 7.80 2.50 508. 1.08 82.7 39.6 1.23 38.3 5.92 7.9 1157. SILTY SAND 4.20 3293. 7.20 21.10 0.00 0.00 6.50 21.10 0.000 2.00 0.789 8.24 2.24 506. 1.11 96.0 40.1 1.30 38.9 6.74 8.5 1176. SILTY SAND 4.40 3812. 8.64 24.90 0.00 0.00 7.83 24.90 0.000 2.00 0.829 9.45 2.18 592. 1.25 109.3 40.2 1.36 39.1 9.04 10.9 1450. SILTY SAND 4.60 2936. 6.50 20.30 0.00 0.00 5.81 20.30 0.000 2.00 0.868 6.69 2.49 503. 0.95 86.5 39.3 1.42 38.2 5.29 6.1 1075. SILTY SAND 4.80 2915. 5.05 14.70 0.00 0.00 4.57 14.70 0.000 2.00 0.907 5.03 2.22 352. 0.73 91.6 39.9 1.49 39.0 3.26 3.6 657. SILTY SAND 5.00 3405. 9.50 28.50 0.00 0.00 8.55 28.50 0.010 2.15 0.938 9.10 2.34 692. 1.27 91.9 38.5 1.52 37.5 10.25 10.9 1671. SILTY SAND 5.20 3211. 11.29 22.25 0.00 0.00 10.74 22.25 0.029 2.10 0.960 11.16 1.07 399. 1.97 14.03 14.6 1040. SILT 5.40 3130. 10.70 23.70 0.00 0.00 10.05 23.70 0.049 2.10 0.982 10.19 1.36 474. 1.44 76.5 36.8 1.57 35.9 13.98 14.2 1192. SANDY SILT 5.40 2630. 9.80 15.40 0.00 0.00 9.52 15.40 0.049 1.95 0.982 9.65 0.62 204. 1.80 11.43 11.6 503. CLAYEY SILT 5.80 3517. 8.55 15.90 0.00 0.00 8.18 15.90 0.088 1.95 1.019 7.94 0.95 268. 1.59 8.76 8.6 608. SILT 6.00 3476. 13.30 25.70 0.00 0.00 12.68 25.70 0.108 2.10 1.039 12.10 1.04 452. 2.07 17.22 16.6 1211. SILT 6.20 3894. 8.80 22.29 0.00 0.00 8.13 22.29 0.128 1.95 1.059 7.55 1.77 492. 1.05 112.9 39.4 1.73 38.7 7.94 7.5 1098. SANDY SILT 6.40 3242. 7.15 20.51 0.00 0.00 6.48 20.51 0.147 2.00 1.078 5.87 2.21 487. 0.87 96.5 38.8 1.75 38.1 5.35 5.0 979. SILTY SAND 6.60 3456. 7.75 19.90 0.00 0.00 7.14 19.90 0.167 2.00 1.098 6.35 1.83 443. 0.92 101.6 38.9 1.79 38.2 6.22 5.7 917. SILTY SAND 6.80 3109. 5.90 15.40 0.00 0.00 5.42 15.40 0.186 2.00 1.118 4.69 1.90 346. 0.72 96.7 39.0 1.82 38.4 3.76 3.4 619. SILTY SAND 7.00 3140. 10.80 23.40 0.00 0.00 10.17 23.40 0.206 2.10 1.138 8.75 1.33 459. 1.28 78.1 36.3 1.81 35.6 12.50 11.0 1088. SANDY SILT 7.20 2977. 16.00 34.00 0.00 0.00 15.10 34.00 0.226 2.10 1.160 12.82 1.27 656. 1.84 55.6 33.1 1.79 32.4 28.75 24.8 1794. SANDY SILT 7.40 3374. 19.30 34.00 0.00 0.00 18.56 34.00 0.245 2.10 1.181 15.51 0.84 536. 2.40 28.84 24.4 1561. CLAYEY SILT 7.60 4159. 17.10 39.90 0.00 0.00 15.96 39.90 0.265 2.10 1.203 13.05 1.53 831. 1.80 92.0 36.1 1.91 35.4 27.92 23.2 2286. SANDY SILT 7.80 4597. 10.50 27.90 0.00 0.00 9.63 27.90 0.285 2.15 1.225 7.63 1.95 634. 1.06 133.5 39.5 2.00 39.0 9.31 7.6 1424. SILTY SAND 8.20 4200. 11.80 24.90 0.00 0.00 11.14 24.90 0.324 2.10 1.269 8.53 1.27 477. 1.21 113.0 38.0 2.05 37.6 12.52 9.9 1119. SANDY SILT 8.20 3221. 8.50 21.50 0.00 0.00 7.85 21.50 0.324 2.00 1.269 5.93 1.81 474. 0.91 91.7 37.4 2.04 36.9 6.81 5.4 950. SILTY SAND***************************************************************************************************************************************************************************
365
Depth 0.6m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 1.2m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 2.1m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 3.3m
0100
200300
400500
600700
800900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 5.1m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)Depth 7.5m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Figure E-1. PMT curve-fitting for PMT1, Texas A&M University.
366
Depth 0.6m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 1.2m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 2.1m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 3.3m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 5.1m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Depth 7.5m
0
100
200
300
400
500
600
700
800
900
1000
0 5 10 15 20 25 30
∆R/R0 (%)
Pres
sure
(kPa
)
Figure E-2. PMT curve-fitting for PMT2, Texas A&M University.
367
Table E-4. Soil Properties from CPT data reduction, Power Plant Mat Foundation
University of Florida
Operator :L. R. M CPT Date :03-29-02 20:45 On Site Loc:CPT Data Cone Used :156 Job No. :CPT01 Water table (meters) : 12.2 Tot. Unit Wt. (avg) : 18.5 kN/m^3----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 0.25 0.82 1.44 11.97 0.83 0.02 sandy silt to clayey silt UNDFND UNDFD 6 .9 0.50 1.64 3.83 31.93 0.83 0.07 silty sand to sandy silt 70-80 >48 13 UNDEFINED 0.75 2.46 6.23 133.30 2.14 0.12 silty sand to sandy silt 80-90 >48 21 UNDEFINED 1.00 3.28 8.62 435.83 5.06 0.16 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 1.25 4.10 10.82 478.93 4.43 0.21 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 1.50 4.92 9.43 478.93 5.08 0.25 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 1.75 5.74 6.85 478.93 6.99 0.30 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 2.00 6.56 6.43 478.93 7.45 0.35 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 2.25 7.38 12.45 478.93 3.85 0.39 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 2.50 8.20 15.39 478.93 3.11 0.44 sandy silt to clayey silt UNDFND UNDFD >50 10.2 2.75 9.02 13.15 478.93 3.64 0.49 sandy silt to clayey silt UNDFND UNDFD >50 8.7 3.00 9.84 10.92 478.93 4.39 0.53 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 3.25 10.66 9.90 478.93 4.84 0.58 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 3.50 11.48 9.84 478.93 4.87 0.62 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 3.75 12.30 7.97 478.93 6.01 0.67 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 4.00 13.12 10.95 478.93 4.38 0.72 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 4.25 13.94 8.72 478.93 5.49 0.76 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 4.50 14.76 20.35 478.93 2.35 0.81 silty sand to sandy silt >90 44-46 >50 UNDEFINED 4.75 15.58 34.16 478.93 1.40 0.86 sand >90 46-48 >50 UNDEFINED 5.00 16.40 34.38 478.93 1.39 0.90 sand >90 46-48 >50 UNDEFINED 5.25 17.22 37.30 478.93 1.28 0.95 sand >90 46-48 >50 UNDEFINED 5.50 18.04 34.48 478.93 1.39 0.99 sand >90 46-48 >50 UNDEFINED 5.75 18.86 31.93 478.93 1.50 1.04 sand to silty sand >90 44-46 >50 UNDEFINED 6.00 19.69 31.40 478.93 1.53 1.09 sand to silty sand >90 44-46 >50 UNDEFINED 6.25 20.51 40.44 478.93 1.18 1.13 sand >90 46-48 >50 UNDEFINED 6.50 21.33 50.13 478.93 0.96 1.18 gravelly sand to sand >90 46-48 >50 UNDEFINED 6.75 22.15 43.64 478.93 1.10 1.23 sand >90 46-48 >50 UNDEFINED 7.00 22.97 37.82 478.93 1.27 1.27 sand >90 44-46 >50 UNDEFINED 7.25 23.79 41.70 478.93 1.15 1.32 sand >90 44-46 >50 UNDEFINED 7.50 24.61 46.12 478.93 1.04 1.36 sand >90 46-48 >50 UNDEFINED 7.75 25.43 42.67 478.93 1.12 1.41 sand >90 44-46 >50 UNDEFINED
368
Table E-4. continued
----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 8.00 26.25 33.96 478.93 1.41 1.46 sand >90 44-46 >50 UNDEFINED 8.25 27.07 30.65 478.93 1.56 1.50 sand to silty sand >90 44-46 >50 UNDEFINED 8.50 27.89 35.44 478.93 1.35 1.55 sand >90 44-46 >50 UNDEFINED 8.75 28.71 36.19 478.93 1.32 1.60 sand >90 44-46 >50 UNDEFINED 9.00 29.53 30.86 478.93 1.55 1.64 sand to silty sand >90 42-44 >50 UNDEFINED 9.25 30.35 25.54 478.93 1.88 1.69 sand to silty sand 80-90 42-44 >50 UNDEFINED 9.50 31.17 20.83 478.93 2.30 1.73 silty sand to sandy silt 80-90 40-42 >50 UNDEFINED 9.75 31.99 27.14 478.93 1.76 1.78 sand to silty sand 80-90 42-44 >50 UNDEFINED 10.00 32.81 37.12 839.08 2.26 1.83 sand to silty sand >90 44-46 >50 UNDEFINED 10.25 33.63 38.85 766.28 1.97 1.87 sand to silty sand >90 44-46 >50 UNDEFINED 10.50 34.45 28.20 574.71 2.04 1.92 sand to silty sand 80-90 42-44 >50 UNDEFINED 10.75 35.27 17.56 429.12 2.44 1.97 silty sand to sandy silt 70-80 40-42 >50 UNDEFINED 11.00 36.09 9.26 574.71 6.21 2.01 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 11.25 36.91 9.58 747.13 7.80 2.06 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 11.50 37.73 9.58 798.21 8.33 2.10 undefined UNDFND UNDFD UDF UNDEFINED 11.75 38.55 9.58 830.14 8.67 2.15 undefined UNDFND UNDFD UDF UNDEFINED 12.00 39.37 9.58 800.77 8.36 2.20 undefined UNDFND UNDFD UDF UNDEFINED 12.25 40.19 9.96 401.91 4.03 2.24 clayey silt to silty clay UNDFND UNDFD 50 6.4 12.50 41.01 17.24 279.54 1.62 2.27 sand to silty sand 70-80 38-40 43 UNDEFINED 12.75 41.83 26.82 269.76 1.01 2.29 sand 80-90 40-42 >50 UNDEFINED 13.00 42.65 35.70 259.99 0.73 2.32 gravelly sand to sand >90 42-44 >50 UNDEFINED 13.25 43.47 31.93 250.22 0.78 2.34 sand 80-90 42-44 >50 UNDEFINED 13.50 44.29 24.84 240.44 0.97 2.36 sand 80-90 40-42 50 UNDEFINED 13.75 45.11 23.37 230.67 0.99 2.38 sand 70-80 40-42 47 UNDEFINED 14.00 45.93 22.80 220.89 0.97 2.40 sand 70-80 40-42 46 UNDEFINED 14.25 46.75 22.22 211.12 0.95 2.42 sand 70-80 40-42 44 UNDEFINED 14.50 47.57 21.65 201.34 0.93 2.45 sand 70-80 40-42 43 UNDEFINED 14.75 48.39 23.30 195.94 0.84 2.47 sand 70-80 40-42 47 UNDEFINED 15.00 49.21 39.03 218.18 0.56 2.49 gravelly sand to sand >90 42-44 >50 UNDEFINED 15.25 50.03 56.99 244.79 0.43 2.51 gravelly sand to sand >90 44-46 >50 UNDEFINED 15.50 50.85 54.84 271.39 0.49 2.53 gravelly sand to sand >90 44-46 >50 UNDEFINED 15.75 51.67 39.27 298.00 0.76 2.55 gravelly sand to sand >90 42-44 >50 UNDEFINED 16.00 52.49 32.42 324.61 1.00 2.58 sand 80-90 40-42 >50 UNDEFINED 16.25 53.31 51.72 351.21 0.68 2.60 gravelly sand to sand >90 44-46 >50 UNDEFINED 16.50 54.13 69.32 377.82 0.55 2.62 gravelly sand to sand >90 44-46 >50 UNDEFINED 16.75 54.95 30.01 404.43 1.35 2.64 sand 80-90 40-42 >50 UNDEFINED 17.00 55.77 20.11 431.03 2.14 2.66 silty sand to sandy silt 70-80 38-40 >50 UNDEFINED 17.25 56.59 37.04 457.64 1.24 2.68 sand >90 42-44 >50 UNDEFINED 17.50 57.41 40.26 908.89 2.26 2.71 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 17.75 58.23 47.00 1229.23 2.62 2.73 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED
369
Table E-4. continued
----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 18.00 59.06 57.79 830.14 1.44 2.75 sand >90 44-46 >50 UNDEFINED 18.25 59.88 34.96 909.96 2.60 2.77 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 18.50 60.70 11.05 989.77 8.96 2.79 undefined UNDFND UNDFD UDF UNDEFINED 18.75 61.52 9.93 941.87 9.49 2.82 undefined UNDFND UNDFD UDF UNDEFINED 19.00 62.34 8.81 510.86 5.80 2.84 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 19.25 63.16 7.69 239.46 3.11 2.86 sandy silt to clayey silt UNDFND UNDFD 31 4.8 19.50 63.98 6.58 239.46 3.64 2.88 clayey silt to silty clay UNDFND UNDFD 33 4.1 19.75 64.80 5.46 239.46 4.39 2.90 silty clay to clay UNDFND UNDFD 36 3.3 20.00 65.62 4.79 239.46 5.00 2.92 silty clay to clay UNDFND UNDFD 32 2.9 20.25 66.44 4.79 239.46 5.00 2.95 silty clay to clay UNDFND UNDFD 32 2.9 20.50 67.26 4.79 239.46 5.00 2.97 silty clay to clay UNDFND UNDFD 32 2.9 20.75 68.08 4.79 239.46 5.00 2.99 silty clay to clay UNDFND UNDFD 32 2.9 21.00 68.90 4.79 239.46 5.00 3.01 silty clay to clay UNDFND UNDFD 32 2.9 21.25 69.72 4.79 239.46 5.00 3.03 silty clay to clay UNDFND UNDFD 32 2.9 21.50 70.54 4.79 239.46 5.00 3.05 silty clay to clay UNDFND UNDFD 32 2.9 21.75 71.36 4.79 239.46 5.00 3.08 silty clay to clay UNDFND UNDFD 32 2.9 22.00 72.18 4.79 239.46 5.00 3.10 silty clay to clay UNDFND UNDFD 32 2.9 22.25 73.00 4.79 239.46 5.00 3.12 silty clay to clay UNDFND UNDFD 32 2.9 22.50 73.82 4.79 239.46 5.00 3.14 silty clay to clay UNDFND UNDFD 32 2.9 22.75 74.64 4.79 239.46 5.00 3.16 silty clay to clay UNDFND UNDFD 32 2.9 23.00 75.46 6.71 254.17 3.79 3.18 clayey silt to silty clay UNDFND UNDFD 34 4.1 23.25 76.28 14.37 312.97 2.18 3.21 silty sand to sandy silt 60-70 36-38 48 UNDEFINED 23.50 77.10 22.35 374.23 1.67 3.23 sand to silty sand 70-80 38-40 >50 UNDEFINED 23.75 77.92 25.54 435.49 1.70 3.25 sand to silty sand 70-80 38-40 >50 UNDEFINED 24.00 78.74 14.85 496.75 3.35 3.27 sandy silt to clayey silt UNDFND UNDFD >50 9.6 24.25 79.56 11.97 558.01 4.66 3.29 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 24.50 80.38 12.57 619.26 4.93 3.32 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 24.75 81.20 16.96 680.52 4.01 3.34 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 25.00 82.02 21.95 741.78 3.38 3.36 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 25.25 82.84 14.37 478.93 3.33 3.38 sandy silt to clayey silt UNDFND UNDFD >50 9.2 25.50 83.66 4.79 191.57 4.00 3.40 clayey silt to silty clay UNDFND UNDFD 24 2.8 25.75 84.48 4.79 191.57 4.00 3.42 clayey silt to silty clay UNDFND UNDFD 24 2.8 26.00 85.30 7.02 209.96 2.99 3.45 sandy silt to clayey silt UNDFND UNDFD 28 4.3 26.25 86.12 23.41 344.83 1.47 3.47 sand to silty sand 70-80 38-40 >50 UNDEFINED 26.50 86.94 35.77 498.08 1.39 3.49 sand 80-90 40-42 >50 UNDEFINED 26.75 87.76 29.33 651.34 2.22 3.51 sand to silty sand 80-90 40-42 >50 UNDEFINED 27.00 88.58 21.85 804.60 3.68 3.53 sand to clayey sand (*) UNDFND UNDFD >50 UNDEFINED 27.25 89.40 15.81 891.57 5.64 3.55 very stiff fine grained (*) UNDFND UNDFD >50 UNDEFINED 27.50 90.22 18.89 574.71 3.04 3.58 silty sand to sandy silt 60-70 36-38 >50 UNDEFINED 27.75 91.04 23.41 478.93 2.05 3.60 sand to silty sand 70-80 38-40 >50 UNDEFINED
370
Table E-4. continued
----------------------------------------------------------------------------------------------------------------------------------- DEPTH Qc (avg) Fs (avg) Rf (avg) SIGV' SOIL BEHAVIOUR TYPE Eq - Dr PHI SPT Su (meters) (feet) (MN/m^2) (kN/m^2) (%) (bar) (%) deg. N bar----------------------------------------------------------------------------------------------------------------------------------- 28.00 91.86 27.94 478.93 1.71 3.62 sand to silty sand 70-80 38-40 >50 UNDEFINED 28.25 92.68 31.37 478.93 1.53 3.64 sand to silty sand 80-90 40-42 >50 UNDEFINED 28.50 93.50 33.17 478.93 1.44 3.66 sand 80-90 40-42 >50 UNDEFINED 28.75 94.32 34.96 478.93 1.37 3.68 sand 80-90 40-42 >50 UNDEFINED 29.00 95.14 36.76 478.93 1.30 3.71 sand 80-90 40-42 >50 UNDEFINED 29.25 95.96 38.55 478.93 1.24 3.73 sand 80-90 40-42 >50 UNDEFINED 29.50 96.78 33.52 478.93 1.43 3.75 sand 80-90 40-42 >50 UNDEFINED 29.70 97.44 32.66 478.93 1.47 3.77 sand to silty sand 80-90 40-42 >50 UNDEFINED-----------------------------------------------------------------------------------------------------------------------------------
Dr - All sands (Jamiolkowski et al. 1985) PHI - Robertson and Campanella 1983 Su: Nk= 15
(*) overconsolidated or cemented
**** Note: For interpretation purposes the PLOTTED CPT PROFILE should be used with the TABULATED OUTPUT from CPTINTR1 (v 3.04) ****
371
LIST OF REFERENCES
Altaee, A., Fellenius, B. H., Prediction of Settlement for Five footings, GeotechnicalSpecial Publication No. 41, Ed. Jean-Louis Briaud & Robert M. Gibbens, FHWA,ASCE, New York, pp. 206-209, June 1994
American Concrete Institute (ACI), Building Code Requirements for ReinforcedConcrete (ACI 318-89) and Commentary-ACI 318R-89, American ConcreteInstitute, Detroit, 1989
American Society for Testing and Materials (ASTM), Annual Book of ASTM Standards,Vol. 04.08 (Soil and Rock; Dimension Stone; Geosynthetics), ASTM,Philadelphia, 1991
Argyris, J. H., Energy Theorems and Structural Analysis, Butterworth, 1960 (reprintedfrom Aircraft Engineering, 1954-1955)
Baguelin, F., Jezequel, J. F., Shields, D. H., The Pressuremeter and FoundationEngineering, Trans Tech Publications, 1978
Baligh, M., Scott, R. F., Quasi-Static Deep Penetration in Clays, ASCE Journal ofGeotechnical Engineering, Vol. 101, No. 11, pp. 1119-1133, 1975
Ballivy, G., The Pressuremeter and its new Avenues (Le Pressiomètre et ses NouvellesOrientations), A. A. Balkema, Rotterdam, 1995
Bathe, K. J., Finite Element Procedures, Prentice-Hall, Inc., New Jersey, 1995
Bell, F. G., Ground Engineer’s Reference Book, Butterworth and Co., London, 1987
Bellotti, R., Ghionna, V., Jamiolkowski, M., Robertson, P. K., Shear Strength of Sandfrom CPT, Proceedings of the 12th International Conference on Soil Mechanicsand Foundation Engineering, Rio de Janeiro, Vol. 1, pp. 179-184, Balkema,Rotterdam, 1989
Bowles, J. E., Foundation Analysis and Design, The McGraw-Hill Companies, Inc., 2nd
Ed., New York, 1977
Bowles, J. E., Foundation Analysis and Design, The McGraw-Hill Companies, Inc., 5th
Ed., New York, 1996
372
Briaud, J.-L., Predicted and Measured Behavior of Five Spread Footings on Sand,Geotechnical Special Publication No. 41, Ed. Jean-Louis Briaud and Robert M.Gibbens, FHWA, ASCE, New York, 1994
Briaud, J.-L., The Pressuremeter, Balkema, Rotterdam, 1992
Briaud, J.-L., Shields, D. H., A Special Pressuremeter and Pressuremeter Test forPavement Evaluation and Design, Geotechnical Testing Journal, ASTM, Vol. 2,No. 3, pp. 143-151, 1979
Broms, B. B., Flodin, N., History of Soil Penetration Testing, Proceedings of theInternational Syposium on Penetration Testing, ISOPT-1, Orlando, Balkema,Rotterdam, No. 1, pp. 157-220, 1988
Bullock, P., The Dilatometer: Current Test Procedures and Data Interpretation, CivilEngineering Report, University of Florida, Gainesville, 1983
Butler, J. J., Analysis of Energy Measurement Methods of SPT Driving Systems, Masterof Science Thesis in Civil Engineering, Utah State University, Logan, 1997
Clarke, B. G., Pressuremeters in Geotechnical Design, Chapman & Hall, UniversityPress, Cambridge, 1995
Clarke, B. G., Smith, A., Self-boring Pressuremeter Test in Weak Rocks, Constructionand Building Materials, Vol. 6, No. 2, pp. 91-96, 1992
Clayton, C. R. I., SPT Energy Transmission: Theory, Measurement and Significance,Ground Engineering, Vol. 23, No. 10, pp. 35-43, 1990
Clough, R. W., Penzien, J., Dynamics of Structures, McGraw-Hill, New York, 1975
Coduto, D. P., Foundation Design, Prentice-Hall, Inc., New Jersey, 1994
Computational Geotechnics, Short Course on Plaxis, Department of Civil, Environmentaland Architechtural Engineering, University of Colorado, (unpublished),August 1999
D’Appolonia, D. J., D’Appolonia, E., Brissette, R. F., Settlement of Spread Footings onSand, Journal of Soil Mechanics and Foundation Division, ASCE, Vol. 96 (SM2),pp. 754-761, May 1968
Das, B. M., Fundamentals of Soils Dynamics, Elsevier Science Publishing Company,Amsterdam, 1983
Das, B. M., Principles of Geotechnical Engineering, 4th Edition, PWS PublishingCompany, Massachussetts, 1998
373
Davidson, J. L., The Cambridge Self-boring Pressuremeter, Short Course, Department ofCivil Engineering, University of Florida, Florida, (unpublished), 1983
Davidson, J. L., Bloomquist, D. G., A Modern Cone Penetration Testing Vehicle,Proceedings of the ASCE Specialty Conference In Situ ’86: Use of In Situ Testsin Geotechnical Engineering, ASCE, New York, pp. 502-513, 1986
Davidson, J. L., Boghrat, A., Displacements and Strains around Probes in Sand,Proceedings of the ASCE Specialty Conference on Geotechnical Practice inOffshore Engineering, Austin, Texas, pp. 181-203, April 1983
Day, R. A., Potts, D. M., Finite Element Analysis of the Hatfield Wall, ComputerMethods and Advances in Geomechanics, Ed. Beer, Booker, & Carter, Balkema,Rotterdam, 1991
De Borst, R., Vermeer, P. A., Possibilties of Finite Elements for Limit Analysis,Géotechnique, Vol. 34, No. 2, pp. 199-210, 1984
Djoenaidi, W. J., A Compendium of Soil Properties and Correlations, Master of Eng. Sc.Thesis, University of Sidney, Sydney, 1985
Douglas, B. J., Olsen, R. S., Soil Classification Using Electric Cone Penetrometer, ConePenetration Testing and Experience, Ed. G. M. Norris & R. D. Holtz, ASCE, NewYork, 1981
Fyffe, S., Reid, W. M., Summers, J. B., The Push-in Pressuremeter: 5 years OffshoreExperience, The Pressuremeter and Its Marine Applications: Second InternationalSymposium, ASTM STP 950, Ed. J.-L. Briaud and J. M. E. Audibert, AmericanSociety for Testing and Materials, pp. 22-37, May 1986
Galerkin, B. G., Series Solution of some Problems of Elastic Equilibrium of Rods andPlates, (in Russian), Vestn. Inzh. Tech., Vol. 19, pp. 8977-908, 1915
Geotechnical Equipment (GPE, Inc.), Data Reduction DMT Basic Program, GeotechnicalEquipment, Gainesville, Florida, April 1993
Gifford, D. G., Wheeler, J. R., Kraemer, S. R., McKown, A. F., Spread Footing forHighway Bridges, U.S. Department of Transportation, Federal HighwayAdministration, Report No. FHWA/RD-86/185, Virginia, October 1987
Handy, R. L., Realism in Site Exploration: Past, Present, Future, and then Some – AllInclusive, Proceedings, Symposium on Site Exploration in Soft Ground Using InSitu Tecniques, Report FHWA-TS-80-202, Federal Highway Administration,Washington, pp. 239-248, 1980
Hayes, J. A., Comparison of Flat Dilatometer Insitu Test Results with ObservedSettlement of Structures and Earthwork, Proceedings, 39th Canadian GeotechnicalConference, Ontario, August 1986
374
Hicher, P. Y., Michali, A., Utilisation de Pressiomètre pour l’Identification desParamèters d’un Modèle Elastoplastique, The Pressuremeter and its NewAvenues, Gérard Ballivy, Ed., Balkema, Rotterdam, pp. 169-178, 1995
Horikoshi, K., Kato, K., Matsumoto, T., Finite Element Analyses of Statnamic LoadingTest of Pile, 2nd Conference on Statnamic Loading Test, Ed. Kusakabe, Kubawaraand Matsumoto, 2000 Balkema, Rotterdam, pp. 295-302, December 1998
Hortvah, J. S., Estimation of Spred Footing Settlement on a Sand Subgrade, GeotechnicalSpecial Publication No. 41, Ed. Jean-Louis Briaud & Robert M. Gibbens, FHWA,ASCE, New York, pp. 145-148, June 1994
Houlsby, G. T., Clarke, B. G., Wroth, C. P., Analysis of the Unloading of aPressuremeters in Sand, The Pressuremeter and Its Marine Applications: SecondInternational Symposium, ASTM STP 950, Ed. J.-L. Briaud and J. M. E.Audibert, American Society for Testing and Materials, pp. 245-262, May 1986
Hrenikoff, A., Solution of Problems in Elasticity by the Framework Method,Transactions of the ASME, Journal of Applied Mechanics, Vol. 8, pp. 169-175,1941
Hughes, J. M. O., Robertson, P. K., Full Displacement Pressuremeter Testing in Sand,Canadian Geotechnical Journal, Vol. 22, No. 3, pp. 298-307, 1985
Hughes, J. M. O., Wroth, C. P., Windle, D., The Pressuremeter Test in Sands,Géotechnique, Vol. 27, No. 4, pp. 455-477, 1977
Jefferies, M. G., Shuttle, D. A., Disturbance Does not Prevent Obtaining ReliableParameters from SBP Tests, The Pressuremeter and its New Avenues, Ed. GérardBallivy, Balkema, Rotterdam, pp. 177-183, 1995
Jeyapalan, J. K., Boehm, R., Procedures for Predicting Settlements in Sands, Settlementof Shallow Foundations on Cohesionless Soils: Design and Performance,Geotechnical Special Publication No. 5, Ed. William O. Martin, ASCE, NewYork, pp. 1-22, 1986
Konstantinidis, B., Riessen, G. V., Schneider, J. P., Structural Settlements at a MajorPower Plant, Settlement of Shallow Foundations on Cohesionless Soils: Designand Performance, Geotechnical Special Publication No. 5, Ed. William O. Martin,ASCE, New York, pp. 54-73, 1986
Kort, D. A., Steel Sheet Pile Wall in Soft Soil, PhD Dissertation, Delft University Press,Delft, 2002
Kovacs, W. D., Salomone, L. A., Yokel, F. Y., Energy Measurements in the StandardPenetration Test, Building Science Series 135, National Bureau of Standards,Washington, DC, 1981
375
Kulhawy, F. H., Jackson, C. S., Mayne, P. W., First Order Estimation of K0 in Sands andClays, Foundation Engineering: Current Principles and Practices, Ed. F. H.Kulhawy, ASCE, New York, 1989
Kulhawy, F. H., Mayne, P. W., Manual on Estimating Soil Properties for FoundationDesign, Report No. EL-6800, Electric Power Research Institute, Palo Alto,California, 1990
Kulhawy, F. H., Trautmann, C. H., Beech, J. F., O’Rourke, T. D., Mcguire, W., Wood,W. A., Capano, C., Transmission Line Structure Foundations for Uplift-Compression Loading, Report EL-2870, Electric Power Reaserch Institute, PaloAlto, 1983
Lamb, R., SPT Energy Measurement with PDA, Paper presented to the 45th annualGeotechnical Engineering Conference at the University of Minnesota, February1997
Lambe, T. W., Whitman, R. V., Soil Mechanics, John Wiley & Sons, Inc., New York,1969
Lewis, C. L., Analysis of Axial Statnamic Testing by the Segmental Unloading PointMethod, Master’s Thesis, Department of Civil Engineering, University of SouthFlorida, Tampa, 1999
Lubliner, J., Plasticity Theory, Mac Millan Publishing Company, New York, 1990
Lunne, T., Robertson, P. K., Powell, J. J. M., Cone Penetrometer Testing in GeotechnicalPractice, Blackie Academic & Professional, London, 1997
Lysmer, J., Vertical Motion of Rigid Footings, Dept. of Civil Eng., Univ. of MichiganReport to WES Contract Report No. 3-115 under Contract No. DA-22-079-eng-340; PhD Dissertation, University of Michigan, Aug. 1965
Malvern, L. E., Introduction to the Mechanics of a Continuum Medium, Prentice-Hall,Inc., New Jersey, 1969
Marchetti, S., The Flat Dilatometer and its applications to Geotechnical Design, LectureNotes (90 pp) Int. Seminar on DMT held at the Japanese Geotech. Society,Tokyo, February 1999
Marchetti, S., A New Insitu Test for the Measurement of Horizontal Soil Deformability,Proceedings of the conference on insitu measurement of Soil properties, Railegh,North Carolina, Vol. 2, ASCE, June 1975, pp. 255-259, 1975
Marchetti, S., Crapps, D. K., Flat Dilatometer Manual, Internal Report of GPE, Inc., 1981
Marchetti, S., Monaco, P., Totani, G., Calabrese, M., The Flat Dilatometer Test (DMT) inSoil Investigations, Report by the ISSMGE Committee TC16. Proc. Insitu 2001,
376
International Conference on Insitu Measurement of Soil Properties, Bali,Indonesia, May 2001
Marcuson,, W. F., Bieganousky, W. A., SPT and Relative Density in Coarse Sand,Journal of the Geotechnical Engineering Division, ASCE, Vol. 103, GT11,pp. 1295-1309, November 1977
Matsumoto, T., A FEM Analysis of a Statnamic Test on Open-ended Steel Pipe Pile, 2nd
Conference on Statnamic Loading Test, Ed. Kusakabe, Kubawara & Matsumoto,2000 Balkema, Rotterdam, pp. 287-294, 1998
Mayne, P. W., Cavity expansion/Critical-State Theory for Piezocone Penetration in Clays(unpublished: submitted for review), 1990
Mayne, P. W., Determining Preconsolidation Stress and Penetration Pore Pressures fromDMT Contact Pressures, Geotechnical Testing Journal, ASTM, Vol. 10, No. 3,pp. 146-150, September 1987
Mayne, P. W., CPT Indexing of Insitu OCR in Clays, Use of Insitu Tests in GeotechnicalEngineering (GSP 6), Ed. S. P. Clemence, ASCE, New York, pp. 780-789, 1986
Mayne, P. W., Kemper, J. B., Profiling OCR in Stiff Clays by CPT and SPT,Geotechnical testing Journal, ASTM, Vol. 11, No. 2, pp. 139-147, June 1988
Mayne, P. W., Kulhawy, F. H., K0-OCR Relationship in Soil, Journal of the GeotechnicalEngineering Division, ASCE, Vol. 108, No. GT6, pp. 851-872, June 1982
McHenry, D., A Lattice Analogy for the Solution of Plane Stress Problems, Journal ofInstitute of Civil Engineers, Vol. 21, pp. 59-82, 1943
McVay, M. C., Numerical Methods in Geomechanics, Course Notes, Department of Civil& Coastal Engineering, University of Florida, (unpublished), Fall 2000
Meyerhof, G. G., Shallow Foundation, Proceedings Journal of Soil Mechanics andFoundation Design, ASCE, March 1965
Middendorp, P., Bermingham, P., Kuiper, B., Statnamic Load Testing of FoundationPiles. Proc. of the 4th Int. Conference on the Application of Stress-Wave Theoryto Piles, A. A. Balkema, The Hague, The Netherlands, pp. 581-588, 1992
Mullins, G., Garbin, E. J., Lewis, C., Statnamic Testing: University of South FloridaResearch, 2nd Conference on Statnamic Loading Test, Ed. Kusakabe, Kubawaraand Matsumoto, 2000 Balkema, Rotterdam, pp. 117-132, 1998
Naval Facilities Engineering Command (NAVFAC), Soil Mechanics, Design Manual 7.1,Department of the Navy U.S. Government Printing Office, Washington D.C.,1982
377
Newmark, N. M., Numerical Methods of Analysis in Bars, Plates and Elastic Bodies,Numerical Methods in Analysis in Engineering, Ed. L. E. Grinter, Macmillan,1949
Nutt, N. R. F., Houlsby, G. T., Measurement and Interpretation of Shear Modulus in SBPTest in Sands, The Pressuremeter and its New Avenues, Gérard Ballivy, Ed.,Balkema, Rotterdam, pp. 115-122, 1995a
Nutt, N. R. F., Houlsby, G. T., Time Dependent Behaviour of Sand from PressuremeterTest, The Pressuremeter and its New Avenues, Gérard Ballivy, Ed., Balkema,Rotterdam, pp. 95-100, 1995b
Oweis, I. S., Equivalent Linear Model for Predicting Settlements of Sand Bases, Journalof Geotechnical Division, ASCE, December 1979
Pal, O., Modélisation du Comportement Dynamique des Ouvrage grâce à des ElémentsFinis de Haute Précision, Thesis, Université Joseph Fourier Grenoble I, Paris,1998
Palacios, A., The Theory and Measurement of Energy Transfert during StandardPenetration Test Sampling, PhD Thesis, University of Florida, Gainesville, 1977
Peck, R. B., Bazaraa, A. R. S., Discussion of Settlement of Spread Footings on Sand (byD’Appolonia and Brissette), Journal of Soil Mechanics and Foundation Division,ASCE, May 1969
Plaxis Version 7 User’s Guide, Ed. Brinkgreve, R. B. J., Vermeer, P. A., A. A. Balkema,Rotterdam, 1998
Powell, J. J. M., Shields, C. C. H., Field Studies of the Full Displacement Pressuremeterin Clays, The Pressuremeter and its New Avenues, Gérard Ballivy, Ed., Balkema,Rotterdam, pp. 239-246, 1995
Prandtl, L., Uber die Eindringungsfestigkeit (Harte) plastischer Baustoffe und dieFestigkeit von Schneiden (On the penetrating strengths (hardness) of plasticconstruction materials and the strength of cutting edges), Zeit. Angew. Math.Mech, 1, No. 1, pp. 15-20, 1921
Randazzo, A. F., Jones, D. S., The Geology of Florida, University Press of Florida,Gainesville, Florida, 1997
Rayleigh, L., On the Theory of Resonance, Transactions of Royal Society, London,Vol. 161, pp. 77-118, 1870
Reddy, J. N., Introduction to the Finite Element Method, McGraw-Hill, New York, 1993
378
Richardson, L. F., The Approximate Arithmetical Solution by Finite Difference ofPhysical Problems, Transactions of the Royal Society (London), Vol. 210,pp. 307-357, 1910
Richart, F. E., Hall, J. R., Woods, R. D., Vibrations of Soil and Foundations, Prentice-Hall, Inc.New Jersey, 1970
Riggs, C. O., North American Standard Penetration Test Practice, an Essay, Use of insitutest in Geotechnical Engineering, Geotechnical Special Publication, No. 6, pp.949-967, 1986
Ritz, W., Über eine Neue Methode zur Lösung Gewissen Variations—Probleme derMathematischen Physik, Journal für die Reine und Angewandte Mathematik,Vol. 135, pp. 1-61, 1909
Robertson, P. K., Campanella, R. G., Interpretation of Cone Penetration Tests: Part I:Sand, Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 718-733, 1983a
Robertson, P. K., Campanella, R. G., Interpretation of Cone Penetration Tests: Part II:Clay, Canadian Geotechnical Journal, Vol. 20, No. 4, pp. 734-745, 1983b
Robertson ,P. K., Campanella, R. G., Gillespie, D., Greig, J., Use of Piezometer ConeData, Proceedings of the ASCE Specialty Conference In Situ ’86: Use of In SituTests in Geotechnical Engineering, ASCE, New York, pp. 1263-1280, 1986
Salgado, R., Mitchell, J. K., Jamiolkowski, M., Cavity Expansion and PenetrationResistance in Sands, ASCE Journal of Geotechnical Engineering, Vol. 123, No. 4,pp. 344-354, 1997
Sanglerat, G., The Penetrometer and Soil Exploration, Elsevier, Amsterdam, 1972
Saran, S., Analysis and Design of Substructures, A. A. Balkema, Brookfield, 1996
Schanz, T., Bonnier, P. G., Verification of Soil Model with Predicted Behavior of a SheetPile Wall, Cumputer Methods and Advances in Geomechanics, Ed. Yuan,Balkema, Rotterdam, 1997
Schmertmann, J. H., Guidelines for Using the CPT, CPTU and Marchetti DMT forGeotechnical Design, Volume III, DMT Test Methods and Data Reduction, U.S.Department of Transportation, FHWA, Pennsylvania, 1988
Schmertmann, J. H., Dilatometer to Compute Foundation Settlement, Proceedings ofInsitu ’86, ASCE Specialty Conference on Use of Insitu Tests in GeotechnicalEngineering, Virginia Tech., ASCE Geotechnical Special Publication No. 6, pp.303-321, June 1986a
Schmertmann, J. H., Suggested Method for Performing the Flat Dilatometer Test,Geotechnical Testing Journal, Vol. 9, No. 2, pp. 93-101, 1986b
379
Schmertmann, J. H., Guidelines for Cone Penetration Test Performance and Design,Report FHWA – TS – 78 – 209, U.S. Department of Transportation, Washington,1978a
Schmertmann, J. H., Use of the SPT to Measure Dynamic Soil Properties?-Yes, But..!,Dynamic Geotechnical Testing, American Society for Testing and Materials,Special Technical Publications, 654, pp. 341-355, 1978b
Schultz, E., Sherif, G., Prediction of Settlements from Evaluated Settlement Observationsfor Sand, Proceedings, 8th International Conference on Soil Mechanics andFoundation Engineering, Moscow, 1973
Seed, H. B., Tokimatsu, K., Harder, L. F., Chung, R. M., Influence on SPT Procedures inSoil Liquefaction Resistance Evaluation, Journal of Geotechnical Engineering,Vol. 111, No. 12, pp. 1425-1445, 1985
Skempton, A. W., Standard Penetration Test Procedured and the Effects in Sands ofOverburden Pressure, Relative Density, Particle Size, Aging andOverconsolidation, Géotechnique, Vol. 36, No. 3, pp. 425-447, 1986
Sluys, L. J., Wave propagation, Localisation and Dispersion in Softening Solids,Dissertation, Delft University of Technology, 1992
Southwell, R. V., Relaxation Methods in Theoretical Physics, Clarendon Press, Oxford,1946
Spoor, K. B., Standard Penetration Testing: Development of an Energy MeasurementProgram for the Florida Department of Transportation, A Master of EngineeringReport, Gainesville, University of Florida, 1997
Symons, I. F., Little, J. A., McNulty, T. A., Carder, D. R., Williams, S. G. O., Behaviorof a Temporary Anchored Sheet Pile Wall on A1(M) at Hatfield, GroundEngineering Division, Structures Group, Transport and Research Laboratory,Crowthorne, Berkshire, RG11 6AU, 1987
Szechy, K. Varga, L., Foundation Engineering – Soil Exploration and SpreadFoundations, Akademiai Kiado, Budapest, 1978
Terzaghi, K., Theoretical Soil Mechanics, John Wiley & Sons, Inc., New York, 1943
Terzaghi, K., Peck, R. B., Soil Mechanics in Engineering Practice, John Wiley & Sons,Inc., New York, 1967
Timoshenko, S. P., Goodier, J. N., Theory of Elasticity, 3rd Ed., McGraw-Hill, NewYork, 1970
Turner, M., Clough, R. W., Martin, H. H., Topp, L., Stiffness and Deflection Analysis ofComplex Structures, Journal of Aeronautical Science, Vol. 23, pp. 805-823, 1956
380
United States Army Corps of Engineers, Design of Sheet Pile Walls, Engineering andDesign, Engineer Manual, EM 1110-2-2504, Department of the Army,Washington, D.C., 1994
William, O. M., Settlement of Shallow Foundations on Cohesionless Soils: Design andPerformance, Geotechnical Special Publication No. 5. ASCE, New York, 1986
Wissa, A. E. Z., Martin, R. T., Garlanger, J. E., The Piezometer Probe, Proceedings of theASCE Specialty Conference on Insitu Measurement of Soil Properties, Railegh,North Carolina, Vol. 1, pp. 536-545, 1975
Withers, N. J., Howie, J., Hughes, J. M. O., Robertson, P. K., Performance and Analysisof Cone Pressuremeter Tests in Sands, Géotechnique, Vol. 39, No. 3, pp. 433-454,1989
Wolffersdorff, P. A., Verformungsprognosen für Stützkonstruktionen,Habilitationsschrift, Veröffentlichung, des Instituts für Bodenmechanick undFelsmechanik der Universität Karlsruhe, Karlsruhe, 1997
Zienkiewicz, O. C., Taylor, R. L., The Finite Element Method, Volume 1, BasicFormulations and Linear Problems, 4th Edition, McGraw-Hill, U.K., 1991a
Zienkiewicz, O. C., Taylor, R. L., The Finite Element Method, Volume 2, Solid and FluidMechanics, Dynamics and Non-Linearity, 4th Edition, McGraw-Hill, U.K., 1991b
Zuidberg, H. M., Piezocone Penetration Testing – Probe Development, Proceedings ofthe 2nd International Symposium on Penetration Testing, ISOPT-1, Orlando,Specialty Session, No. 13, Balkema, Rotterdam, 1988
381
BIOGRAPHICAL SKETCH
The author was born in January 1968 in Antananarivo, Madagascar; fourth child
but last son of the five children of his parents. He attended the College St. François
Xavier Antanimena, Antananarivo up to Junior High School, then entered the Lycée
Faravohitra (ex-Lycée Jules Ferry) and graduated from High School in 1984. After
completing one year of military service at the Ministry of Defense in Madagascar, he
entered the Polytechnic Insitute of Antananarivo in January 1986. He obtained his degree
DUET in 1988, and then the degree “Ingénieur de Bâtiment et Travaux Publics” in June
1991. He then worked for a building construction company TAOTRANO, and then for
the French construction control agency of SOCOTEC in Madagascar. In 1993, he was
offered a scholarship for graduate studies by the Japanese Government. He earned his
Master’s degree from the Department of Materials Engineering, Miyazaki University,
Miyazaki, Japan, in March 1996. Also there he did theoretical works and numerical
analysis on Homogenization Theory applied to roller-compacted concrete structures
through March 1999. After working for the Japanese general contractor DAIHO
Corporation, he moved on to the University of Florida, U.S.A in August 1999. He began
his advanced education in geotechnical engineering at the University of Florida,
Geotechnical Division, which was his ultimate goal. His long-term professional ambition
is to contribute to the design and construction of various civil engineering structures.
