Correl Soils Properties Carter&Bentley

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Í IT S



c

LAT

SO L

Michaef Cárter

n

Stephen P Bentley

PENTECH

PRESS

Publishers London



ref ce

E n g i n e e r s a n d g e o l o g i s t s a r e o f t e n e x p e c t e d t o

g i v e

p r e d i c t i o n s o f s o il

b e h a v i o u r

e v e n

w h e n

l i t t l e o r n o

r e l e v a n t

t e s t r e s u l t s a r e a v a i l a b l e .

T h i s

i s pa r t icu la r }

7 t r u e

o f s m a l l

p r o j e c t s

o r fo r

p r e l i m i n a r y d e s ig n s .

O u r a i m i n t h i s b o o k h a s b e e n t o g a t h e r t o g e t h e r m a t e r i a l t h a t vvou l d

b e o f

p r a c t i c a

a s s i s t a n c e t o t h o s e f a ced w i t h t h e p r o b l e m o f h a v i n g t o

e s t í m a t e

s o i l b e h a v i o u r

f r o m

l i t t l e o r n o l a b o r a t o r y t e s t d a t a .

T h e f i e l d o f s o i l p r o p e r t y c o r r e l a t i o n s is d i v e r s e a n d c o m p l e x a n d

o u r m a i n d i f f i c u l t y i n

p r o d u c i n g

th e

w o r k w a s

th e

v o l u m e

o f

m a t e r i a l

a v a i l a b l e . C o n s e q u e n t ly , w e h a v e h a d t o b e s e l ec t i v e i n o u r a p p r o a c h

a n d w e

h o p e

t h a t o u r f i n a l c h o í ce p r o v i d e s a w o r k a b le c o m p e n d i u m .

M o d e r n i n - s i t u t e s ti n g m e t h o d s

i s a

r a p i d l y d e v e l o p m g a s p ec t

o f

g e o t e c h n i c a l

e n g i n e e r in g

w h i c h

w a r r a n t s a

t e x t

to i t s e l f : t h i s a s p ec t i s

n o t d e a l t

w i t h h e r e b u t ,

w h e r e

a p p r o p r i a t e , s u i t a b l e r e f e r e n c e s

a r e

g i v e n .

T h e w o r k p r e s e n t s t y p i c a l v a l ú e s o f e n g i n e e r i n g p r o p e r t i e s fo r

v a r i o u s t y p e s

o r

c l asses

o f

s o i l , t o g e t h e r w i t h c o r r e l a t i o n s b e t w e e n

d i f f e r e n t

p r o p e r t ie s . P a r t i c u l a r e m p h a s i s i s g i v e n to c o r r e l a t i o n s w i t h

soi l c l a s s i f í c a t i on

t es t s and t o t he u se o f c l a s s i f i c a t i o n s y s t e m s .

I n c l u d e d i n t h e c o r r e l a t i o n s a r e p r o p e r t i e s t h a t a r e d i f f í cu l t t o

m e a s u r e d i r e c t l y ,

s u c h a s f ros t

s u s c e p t i b i l i t y

an d

s w e l l i n g p o t e n t i a l .

In

a d d i t i o n ,

s o m e

e x p l a n a t i o n s a re g i v e n o f t he e n g i n e e r i n g r e le v a n c e o f

th e v a r i o u s p r o p e r t i e s a n d t h e j u s t i f i c a t i o n o f t h e c o r r e l a t i o n s

be tw

;

een

p r o p e r t i e s i s d i s c u s s e d .

S u c h p r e d i c t i o n s c a n ,

o f

c o u r s e , n e v e r

b e a

s u b s t i t u t e

f o r

p r o p e r

t e s t i n g b u t w e h o p e

t h a t

th e

i n f o r m a t i o n

i n

t h i s b o o k

w i l l

e n a b l e

o p t i m u m u s e o f so i l c l ass i f íca t ion

d a t a .

S t e p h e n P B e n t l e y

C a r d i f f , W a l e s

M i c h a e l C á r t e r

C o l o m b o , S r i L a n k a



Contents

CHAPTER 1 GRADING AND PLASTICITY 1

1.1 GRADING 1

1.1.1 The influence of grading on soil properties 1

1.1.2

Standard grading divisions and sieve sizes 3

1.2

PLASTICITY 3

1.2.1 Consistency Limits 6

1.2.2

Development

of the

l iquid

and

plástic

limit

tests

7

1.2.3

The shrinkage

l imit

test 8

1 2 4 Consistency limits as indicators of soil behaviour 10

1.2.5 Limitations on the use of consistency

limits

12

CHAPTER 2 SOIL CLASSIFICATION SYSTEMS 13

2.1 COMMON SOIL CLASSIFICATION SYSTEMS 14

2.2

CORRELATION

OF THE

UNIFIED

BS AND

AASHTO SYSTEMS 38

CHAPTER

3

DENSITY

39

3.1 NATURAL DENSITY 39

3.2 COMPACTED DENSITY 43

3.2.1 Compaction test standards 43

3 2 2 Typical compacted densities

45

3 2 3 Typical moisture d ensity curves 49

CHAPTER

4

PERMEABILITY

50

4.1 TYPICAL VALÚES 51

4.2 PERMEABILITY AND GRADING 51



CHAPTER CONSOLIDATION AND SETTLEMENT

5 1

5 2

COMPRESSIBILITY

OF

CLAYS

5 1 1 The compressibility parameters

5 1 4

Typical

valúes

and

correlations

of com pressibility

coefiícients

5 1 5 Settlement corrections

RATE OF CONSOLIDATION OF CLAYS

5 3 SECONDARY COMPRESSION

5 4

SETTLEMENT

OF SANDS AND GRAVELS

5 4 1 Probes and standard penetration tests

5 4 2 Píate bearing tests

55

56

5 1 2 Setílement calculations using consolidation theory 58

5 1 3 Settlement calculations using elasticiíy theory 59

9

60

62

65

68

7

7

7

CHAPTER 6 SHEAR STRENGTH 76

6 1

THE CHOICE OF

TOTAL

OR

EFFECTIVE STRESS

ANALYSIS 78

6 1 1 The choice in practice 79

6 2 UNDRAINED SHEAR STRENGTH OF CLAYS 80

6 2 1

Rem oulded shear strength 81

6 2 2 Undisturbed shear strength

83

6 2 3 Predictions using the standard penetration test 89

6 3

DRAINED

AND EFFECTIVE SHEAR

STRENGTH

OF

CLAYS

89

6 4

SHEAR

STRENGTH

OF G RA N U LA R

SOILS

90

6 5 LATE RAL PRESSUR ES IN A

SOIL MASS

92

CHAPTER

CALIFORNIA BEA RING RATIO 97

7 1 THE TEST METHOD 97

7 2

CORRELATIONS WITH SOIL CLASSIFICATION

SYSTEMS

97

7 3 CBR AND

SHEAR

STRENGTH 104



CHAPTER

S H RI N K A G E

AND

SW ELLING

CHARACTERISTICS

8 1 IDENTIFICATION

8 2 SWELLING P O T E N T I A L

8 2 1 Relat ion

to

o the r

proper t ies

8 3 S W ELLI N G P RES S U RE

5

5

107

107

113

CHAPTER

9

FROST SUSCEPTIBILITY

9 1 ICE S EG REG A TI O N

9 2 G R A I N S I Z E S

9 3

PLASTICITY

e f e r e n c e s

n d e x

7

9

8



Chapter

GRADING

AND PLASTICITY

The concepta of grading and

p lasticity

and the use of

these properties

to iden tify classify and

assess

soils are the

oldest

and

most

fundamental in

soil mechanics. Their use

in fact pre-dates th e

concept

of

soil mechanics

itself: th e

basic ideas w ere borrow ed from

pedologists

and soil scientists by the fírst soil engin eers as a basis for

their new

science.

1 1

GRADING

It can be

readily appreciated

by

even

th e

most untrained

eye

t ha t

grave l is a

somewhat

diíferent

material from sand. Likewise

silt and

clay are different

again. Perhaps

not

quite

so

obvious

is

that

it is not

just

th e

particle size tha t

is

impor tant

bu t the distribution of

sizes th at

make up a

particu lar soil. Thus

the

grading

of a

soil determines ma ny

of its characteristics. Since it is such an o bviou s property and easy to

measure

it is

plainly

a

suitable fírst choice

as the

most fundam ental

pro perty to assess the characteristics of soil at least for coarse grained

soils.

Of

course

to

rely

on

grading alone

is to

overlook

th e

influences

of

such characteristics as particle shape mineral comp osition and

degree

of compaction.

Nevertheless grading

has been found to be a

major

factor

in determining the

properties

of

soils particularly

coarse-grained soils w here min eral compo sition

is

relatively

unim

portant.

1 1 1 The influence of grading on soil properties

During th e

early

development of

soil mechanics engineers relied

heavily on past experience and

found

it

convenient

to classify

soils

so

that experience gained

w i th a

particular type

of

soil could

be

used

to

assess the suitability of similar soils for any specific purpose and to

indícate appropriate methods of treatm ent. Thus the concept of soil

classification aróse early in the development of soil mech anics. Ev en



2 CORRELATIONS O F SOIL PROPER TIES

today, despite

the

development

in

analytical techniques

which has

taken place, geotechnical engineers rely heavily on past experience,

and soil

classification

sys tems are an inva luab le aid , part icu larly

where soils are to be used in a remoulded form,

such

as in the

construct ion of e m b a n k m e n t s and filis. The use of grading in soil

classifíations

is

discussed

in

Chapter

2 .

Poorly-graded soils, typically

trióse with a

very small range

of

particle sizes, con tain

a higher

proport ion

of

voids than w ell-graded

soils,

in

wh ich

the fíner

particles

fíll the

voids between

the

coarser

grains. T hu s, grad ing iníluences the d ens ity of soils. This is indic ated

in a general w ay in Ch apter 3 Table 3.1). An oth er consequ ence of the

greater degree of packing achievable by well-graded soils is that the

proport ion

of

voids w i th in

the

soils

i s

reduced.

In

addit ion, al though

th e

proportion

of

voids

in fine-grained

soils

is

relative ly high,

the

size

of

individual voids is ex trem ely small. Since the proportion and size of

voids aíTecí íhe flow of water th rough a soil, grading can be seen ío

influence permea bil ity . The theoret ical relat ionship between grading

and

permeabil i ty

is

discussed

in

Chapter

4 and the

coefficient

of

permeabil i ty

is

related

to

grain size

in

Figure 4.1.

Since consolidation

involves the

squeezing-out

of

wa te r

from the

soil

voids, as the soil grains pack closer to geth er un de r load, it follow s

that

th e

rate

at

w hich consolidation takes place

is

controlled

by the

soil permeability. Since permeability

is, in

turn, partly controlled

by

grading

it can be

seen that grading

influences th e

rate

of

consolida-

tion.

Also,

since fíne-grained

soils

and

poorly-graded soils ha ve

a

higher proportion

of

voids,

and

tend

to be less

well-packed than

coarse-grained

and

well-graded

soils

they tend

to consolídate

more.

Thus the

consolidation properties

of a soil are

profoundly

iníluenced

by

its

grading. Since

fine-grained

soils tend

by and large to be

more

compressible

than

coarse-grained

soils

and consolídate at a much

slower rate

it is

these soils that

are of

most concern

to the

engineer.

Their gradings

are

much

too fine to be

measured

by

conventional

means and,

at

these sm all particle sizes,

th e

properties

of the

minerals

present are of more importance than th e grading. Specific correla-

tions between grading

and

consolidation chara cteristics

do

not,

therefore, exist. However,

th e

efíect

of

grading

on

consolidation

is

taken into account indirectly in some soil

classifications

which are

used

to

assess

th e

suitability

of

soils

for

earthworks

and

pavement

subgrades.

Shear strength is also

affected

by grading since grading influences

th e amoun t of interlock between particles bu t correlations between

grading and shear strength are not possible because other factors,

such as the angularity of the

particles

th e confíning pressure, th e



G R A D I N G A N D PLA STIC I TY 3

compaction

and

consolidat ion history ,

and the

types

of the clay

minerals

are of

overriding importance.

The

variability

of

some

of

tríese

factors

is

reduced where

only

compacted

soils are

considered

and, with the aid of

soil

classifícation system s, the

iníluence

of grading

on shear stren gth can be given in a general way, as indicated in Table

6.2. Similarly,

the

influence

of the

g rad ing

o f

coarse-grained soils

on

their California bearing ratio

is

ind icated

in

Table

7 .2

an d ,

to

some

extent , in

Figure

7.3.

In a broad

sense, both swelling properties

and

frost susceptibility

are influenced by grading. Correlation between grain size and

frost

susceptibility

can be

seen

in

Chapter

9 but the

identifícation

of

expansive

clays, discussed

in

Chapte r

8,

relies alm ost

entirely on the

plast ici ty

properties, the only re levan t aspect of grading being the

propor t ion

of

material

finer

than 2/rni.

1 1 2

tandard grading

divisions and

sieve sizes

Although

grading, as the mo st basic of soil propertie s, is used to bo th

identify and classify

soils,

th e

división

of

soils into categories, based

on grading,

varíes

according to the agency or classifíca tion system

used.

A comparison of

some common

defínitions

used

is

given

in

Figure

1.1.

For

soil particles larger than

60¿on,

grading

is

carried

out

using

standa rd square mesh sieves. Table

1.1

shows s tandard

sieve

sizes

and

gives a

comparison between British

and

American standards.

1 2

PLASTICITY

Just

as the

concepts

of particle size and grading can be readily

appreciated for coarse-grained soils, so it is obvious that clays

are

somehow fundamenta l ly different

from

coars e-grained soils, since

clays exhibit the property of plasticity whereas sands and gravéis do

not.

Plasticity is the

ability

of a

material

to be

mou lded irreversibly

deformed)

w i thou t

fracturing.

In

soils,

it is du e to the

electrochemical

behaviour

o f the

clay minerals

and is

un ique

to

soils containing

clay-

mineral particles.

These are plate-like

structures which typically

possess a negative electrical charge on their

face

surface, brought

about by inherent flaws within the chemical lattice. In nature, this

negative

charge

is

cancelled

out by cations Na + ,

Ca+

+

etc.) present

in

the pore

water.

The

positive

to

negative attraction, between

the



CORRELATIONS OF

SOIL

PROPERTIES

British Standard and MIT

clay

silt

m

c

sand

m

c

grave

m

c

cobb-

les

boulders

O OO2

O O O 6

O O2 O O6 0 2 0 6

6 2O 6O 20O

Unif ied Soil Classif¡catión System

fines silt, clay )

sand

m ] c

gravei

f

c

cobb-

les

bouiders

0.075 0.425 2 4.75

19

75

300

AST1KD422,

D653)

fines

silt,

clay

)

sand

f | m|

gravei

«Ato-

les

bouiders

O 075

0 425 2

4 75

75 300

AASHTO T88)

colloids

clay

silt

sand

f

c

gravei

bouiders

O CO 1 O O O 5

O 075 0 425

Grain size

)

LL

1 I 1 lu. S i l . |t lI I 1 lu. I it i InnI

75

_ÍL1.1_1_5

luí

i l i i

i

0.001

O.01

0.1

10

100 10OO

Figure

1.1 Some

common

dejlnitions ofsoils, classijled by

par ticle size

modified after

Al-Hussaini ,

1977)

catión and the clay mineral, pro vides a netw ork of bonds throu gho ut

the clay mass, as illustrated in Figure 1 2 Also,

because

water

molecules themselves are polarised, water molecules immediately

adjacent to the clay minerals become attracted and

bonded

(adsor-

bed) to the surface to form an adsorption com plex . Since these

electrochemical bonds act

through

the water surrounding the clay

particles,

th e

at traction

is

maintained even w hen

large deformations

take place between clay particles,

to

produce

the phe

orne ion

of

plasticity.

Plástic

soils - clays - are often described as cohesive to distmguish

them from

non -plastic

soils

-

sands and gravéis

-

which

are

described

as

granular

or

non-cohesive . Thus,

th e terms plástic and cohe-

sive are often

used synonymously. Since

all

plástic soils

a re

cohesive

and all coh esive soils are plástic this

seems

quite reasonable,

yet,

not



G R A D I N G A N D PL A ST ICIT Y

Table 1 1 C O M P R I S O N

O F

S T N D R D

S I E V E S

T Y P I C L L Y U S E D

I N

S O I L T E S T I N G

Aperíure

size

75mm

63mm

50mm

37.5mm

28ram

25mm

20mm

19mm

14mm

12.5mm

lO.Omm

9.5mm

6.3mm

S.Omm

4.75mm

3.35mm

3.18mm

2.36mm

2.00mm

1.70mm

l.ISmm

850/mi

600^m

425/^m

300/zm

250/im

150¿un

75/im

63/ím

These sieve sizes are

_2 M0 «

Í/.S.

sieve

designation

3in

2în

2in

l|in

l in

lin

U n

f in

¿in

No. 4

No. 8

No 16

No. 20

No. 30

No. 40

No. 50

No. 60

No. 100

No. 200

either unavailable or

•ww? *

B.S. sieve

designation

75mm

63mm

50mm

37.5mm

28m

20mm

14mm

lOmm

6.3mm

5mm

3,35mm

2.00mm

1.70mm

1.18mm

850/im

600/zm

425/im

300/im

100/zm

75/zm

63/ím

are not normally

used.

Oíd

Imperial)

B S sieve

designation

3in

2iin

2in

lîn

l in

|in

lin

f in

¿in

16

sin

No. 7

No. 10

No. 14

No. 18

No. 25

No. 36

No. 52

No. 60

No. 100

No. 200

„

v L

1

a b

Figure

1.2 Electrochemical bonding

between clay-mineral

par

fieles;

a) dispersed

structure

b) flocculated síructure



6 C O R R E L A T I O N S

OF

SOIL P R O P E R T I E S

only are the tw o prope rties

subtly

diíferent in nature , their underlying

cause is quite different. Whereas plasticity is the property that allows

deformation w i thou t cracking, cohesión is the possession of shear

strength which allows

the soil to

maintain

it s

shape under load even

when it is not confíned. And wh ereas plast ici ty is produced by the

electrochemical nature of the

clay

particles, cohesión occurs as a

result

of

their very

small

size, which results

in

extremely

low

permeabilit ies

and

al lows pore water pressure changes during

defo rma tion tha t gives clays the shear stre ngth prope rties w e describe

as cohesive. The precise mechanism involved is described more

thoroughly in

Chap ter

6, but three

simple examples help il lustrate

these d iíferences. Firstly, althou gh sands

cannot be

moulded wi thou t

cracking, they

can

possess

a

weak cohesión, al lowing children

to

m a k e sandpies and sandcastles. This is actua lly the result of m enisc us

forces

in

partially-saturated sands,

and disappears in saturated

condit ions, Secondly, if clays

are

loaded sufficiently síowly, íheir

strength characteristics are similar to those of granular soils; tha t is ,

they

behave

like frictional

materials . Again, this

is

discussed more

fully in

Chapter

6.

Thirdly,

non -plast ic silts,

which

are

composed

of

very

small

particles of un altered

rock,

do possess a

transient cohesión,

even thou gh they are non -plastic. Thus, it can be seen tha t plasticity

and cohesión go

together

not because the y are different facets of the

same property

but

because clay particles

are at the

same time both

extremely small and co mposed of minerals, the

producís

of chem ical

alteration

that

possess

particu lar electrochemical feature s.

1 2 1 onsistency l imit s

The notio n of soil consistency limits stems from the concept tha t soil

can exist in an y of

four

states, dep end ing on its mois ture conten t. This

is illustrated

in Figure

1.3,

where

soil

is

shown settling

out of a

suspensión in water, and

slowly

dr yin g ou t. Initially, the soil is in the

form of a viscous liquid,

with

no shear strength. As its mois tu re

content is reduced , it begins to attain som e strength but is still easily

moulded: this is the plastic-solid phase. Further drying reduces its

ability to be m oulded so that it tends to crack as

m ould ing

occurs: this

is the sem i-solid

phase. Eventual ly,

th e

soil becom es

so dry

tha t

it is a

brittle

solid. Early

ideas

on the

co nsistency concept

and

procedures

fo r

its measurement were

developed

by

Atterberg,

a

Swedish chemist

and

agricultural researcher

in

about 1910.

In his

original work

Atterberg 1911)

identifíed fíve

limits

bu t

only three

shrinkage,

plástic

and

liquid limits) have been used

in

soil m echanics.

The

liquid

and plástic limits represent the moisture contents at the borderline



G R A D I N G A N D P LA S TI CI TY 7

' ''. ' .'

• -

• • . : •

; •

llfi?

Liquid Viscous

suspensión liquid

' / ' ' / ^

(' T V/-t

Plástic

solid

S

£M^¡

emi-plastic

solid

%%®® &,

Solid

a

u

m

Solid

< n

•

O

w

E

1

- Plástic

a

•• •

=Liquid

•o

W at e r ontent

b )

Figure 1.3 Consistency

limits:

o)

change

from liquid to solid as a soil dries out b)

volume and

consistency

changes wiíh

water content change

between plástic and l iquid phases and between semi-sol id and solid

phases,

as

indicated

in

Figure 1 3

The

shrinkage l imit represents

th e

moisture content at which

fur ther

dry ing of the soi l causes no

fu r the r

reduction

in

volume. This

is

illustrated

ín

Figure

1.3 b) . In

elec-

t rochemical

terms, the clay mineral part ic les are far enoug h apa rt a t

the l iquid l imit

to

reduce

the

elect rochemical a t t ract ion

to

a lmost

zero, and at the plást ic l imit there is the minimum amount of water

present to maintain the flexibility of the bonds.

1.2.2 Development of the liquid and plástic l imit

tests

The methods of measurement of the l iquid and plást ic l imits have

changed Hule since 1910 The me t hod of hand-rolling clay into fine

threads to determine the plást ic l imit remained virtual ly as i t was

originally

defined

unti l H arison 1988) suggested

a

procedure using

a

cone penet rom eter .

The

liq uid limit test,

in

wh ich soil

w as

originally

held

in a

cupped

hand and

tapped gently, evolved

to

provide



8 CORRELATIONS OF SOIL PROPERTIES

Table 1 2

C O R R E C T I O N F A C T O R S F O R T H E

O N E - P O I N T

L IQUID

L J M I T

T E S T

No of

blows

15

16

17

18

19

20

21

Factor

F

0 9 5

0.96

0.96

0.97

0 9 7

0.98

0.98

No of

blows

22

23

24

25

26

27

28

Factor

F

0.99

0.99

0.99

1 0 0

1.00

1.01

1.01

No of

blows

29

30

31

32

33

34

35

Factor

F

1 0 1

1.02

1.02

1.02

1 0 2

1.03

1.03

Liquid limit = moisture contení of test specimen x factor F .

much-needed standardisation:

a

metal

dish replaced the cupped hand

and the Casagrande apparatus, developed in 1932, replaced the

original hand-tapping.

The

introduction

of the

cone penetrometer

method

in 1922 fur ther improved

repeatability

of the

liquid limit test.

When th e Casagrande method is used to determine th e liquid l imi t ,

a

plot

is drawn of moisture coníent against blow count (to a

logarithmic scale). For

soils

of a similar geológica origin, the slope of

the plot is

similar,

so

that once

one

point

has

been established,

it is

possible

to

draw

a line

through

it, at the

correct slope

to

obtain

an

approximate valué

of the

liquid limit

w i t h o u t the

need

fo r furíher

testing:

this

is the

one-point Liquid Limit test.

All

British

soils have

been

found

to show a similar slope so that their liquid limits

m a y

be

obtained

in this way. As an alternative to constructing a

graph,

liquid

limit valúes are obtained by multiplying the moisture

contení

valué of

the

test specimen

by

a

correction

factor, obtained from Table

1.2.

Results a re less accurate than for the

full

test procedure but tesing i s

much quicker.

1 2 3 The shrinkage limit test

The

shrinkage limit test

is difíicult to

carry

out and

results vary

according

to the

test method used

¿ nd

sometímes even deoend

on the

initial moisture

contení of the

test specimen.

If íhe

specimen

is

síowly

dried

from a

water

contení

near

the auid

limit (for

exarr de,

using

the ASTM D 427 procedure), a shrinkage limit valué of giv ,ter than

th e plástic limit m ay be obtained; this is meaningless when considered

in the contexí of Figure 1.3. This is paríicularly írue wiíh sandy and

silíy

clays. Likewise,

if íhe

soil

is in iís

naíural, undisíurbed

síaíe

íhen

the

shrinkage

l imií is often

greater

íhan the

plástic limit

due to the

soil

structure

(Holíz and

Kovacs

1981).

Karlsson (1977),

who

carried

out



GRA DING AND PLASTICITY 9

shrinkage

limit

tests

on a

n u m b e r

of

Swedish clays,

found

t ha t

shrinkage

l im it was related to sensit ivity

discussed

in

C hap te r

6). For

clays of médium sensitivity the shrinkage limit of undisturbed

samples was about equal to the plástic l imit , whereas undisturbed

highly sensitive clays showed shrinkage limits greater

than

the

plástic

limits. Un disturbed organic clays showed sh rinka ge l imits

well

below

th e

plástic limits.

For

all

th e soils

tested,

th e

shrinkage l imits

o f the

disturbed sam ples were lower tha n thos e of the undis turbe d samples,

and below the plástic limit.

In his lectures at Harvard University, Casagrande suggested that

the ini t ia l moisture con ten í for sh rinkag e l im it tests should be slightly

above the plástic lim it, but it is

difficult

to prepare specimens to such

low moisture contents without entrapping air bubbles. It has been

found tha t for soils prepared in this way and tha t plot near the

A-line

of

a

plasticity

chart

see Figure 2.1 ,

the

shrinkage l imit

is

about

20.

If

the soil plots a n

a m o u n t

A p

vert ically abov e

o r

below

the

A-line, then

the

shrinkage limit

will be less

than

or

greater than

20 by A p.

That

is

fo r

plots

Ap

above

the

A-line

=

20-Ap

Soil B SL = 7

Soil A SL = 4

Figure

1.4 Casagrande s procedure fo r estimating th e

shrinkage limit



10

CORRELATIONS

OF SOIL

PROPERTIES

For ploís

p

below íhe A-line

This procedure

ío

deíermine

íhe

shrinkage

limií

(for

soils

prepared

in

the

manner suggested

by

Casagrande)

has

been

found

ío be as

accuraíe

as íhe íesí itself. An alternaíive and even simpler procedure is

illusíraíed

in Figure

1

.4. The U-line and A-line of íhe plasíiciíy

charl

are

exíended

ío meel ai

co-ordinaíes

43.5,

—46.4) and a line is

drawn

from íhe

ploííed poiní

ío

íhis

inlerseclion, as

illusíraíed. This

line crosses

íhe

liquid limií axis

ai a

valué approximaíely

equal ío íhe

shrinkage

limit.

1.2.4 Consistency limits as indicators of soil behaviour

The

liquid limit should, from

the way it is defined in

Figure

1

.3 ,

be íhe

minimum

moisture contení

ai

which

íhe

shear sírengíh

of the

soil

is

zero.

However, because

of the w ay the

standard liquid limit tesis have

been defíned, the soil actually has a

small

shear sírength. The

Casagrande procedure models a slope

failure

due ío dynamic loading

under quick undrained condiíions. The shear strengíh of the speci-

men is progressively reduced b y increasing iís moisíure

conlení

until

a

speciííc

energy inpuí, in íhe form of síandard íaps, causes a failure of

a standard

slope

in íhe

defíned manner.

The

alíernative cone method

devised

by íhe

Swedish Geotechnical Commission

in

1922,

is

also

an

indirecí shear sírengíh test thaí models

bearing

failure

under quick

undrained condiíions.

The

consequence

of

these

tesl

procedures

is

that

all

soils

at

their liquid limil exhibit

íhe

same valué

of

undrained

shear sírengíh. Casagrande

(1932)

eslimaled this

valué

as 2.6kN/m2

and laler work by Skemplon and Norlhey (1952) indicated valúes of

l-2kN/m2.

The hand

rolling

procedure used in íhe plasíic limil

lest

can be regarded as a measure of the toughness of a soil (íhe energy

required ío fracíure

il )

which is also relaled lo

shear

sírengíh,

although

there

are n o

obvious analogies

for íhe

mechanism

of failure.

Il has

been

found Ihat all

soils

at the

plástic limit exhibit similar valúes

of

undrained shear strengíh reported

by a

number

of

researchers

as

being

100-200kN/m2. Il was

recognised

as early as

1910

Ihal íhe

consislency limil

lesls

are measures of shear strengíh, and Atlerberg s

assislanl íhe geologisl Simón Johansson, presenled

an

árdele

on

íhe

sírengíh

of soils al

different

moisíure conlenls in 1914.

From íhe preceding discussion

il can be

seen Ihaí

all

remoulded

soils change íheir sírengíh

Ihroughoul

Iheir plasíic range from aboul

IkN/m2

al íhe liquid limil lo abouí 100kN/m2

al

the plástic limit. The

plasticiíy

índex is Iherefore íhe

change

of

waíer conlení needed

lo

bring

aboul

a

sírengíh change

of

roughly

one hun dred-fold,

within



G R A D IN G AND PLASTICITY 11

the plástic range of the soil. A remoulded soil with a mo isture content

within the plástic range can be expected to have a shear strength

somewhere between these extremes and it seems

reasonable

to

assume that , for a giv en soil, it s actu al shear strength will be related to

its moisture con tent.

Also,

assuming th at the general pa ttern of shear

strength change with moisture content, across the plástic range, is

similar for all

soils,

then i t

should

be possible to predict th e remoulded

shear strength of any clay

from

a knowledge of its m oisture content

and its

liquid

and

plástic limits. Correlations

of

remoulded shear

strength an d m oisture con tent, related to the liquid and plástic

l imit,

have been obtained and are discussed in Chapter 6. With

slight

corrections

and some loss of

accuracy , these co rrelations

may

also

be

used to predict th e shear strength of undis turbed

clays.

This is

especially useful in view of the

fací that most

clays, both in their

natural state and when used in earth w ork s, are in a plástic state.

A further consequence of these concepts is that a soil with a low

plasticity

Índex

requires only a small reduction in mo isture content to

bring about a substan tial increase in shear streng th. Con versely, a soil

with a high plasticity Índex will not stabilise under load until large

moisture content changes have taken place. This implies that highly

plástic soils will

be

less stable

and

that

a

correlation

may

exist

between p lasticity

and

com pressibility. Also,

the

liquid limit d epends

on the

amounts

and

types

of

clay m inerals

present

which

control the

permeability, henee the rate of consolidation, imp lying a c orrelation

between

liquid limit

and the

coefíicient

of

consolidation. Consolida-

tion properties are discussed in Chapter 5.

The

special

property of

plasticity

in clays is a function of the

electrochemical behaviour

of the clay mine rals: soils tha t possess no

clay m inerals do not exhib it plasticity and, as their moisture content

is reduced, they pass directly from the liquid to the

semi-solid

state.

The Atterberg limits can give indications of both the type of clay

minerals present and the amount . The ratio of the p lasticity Índex to

the

percentage

of

m aterial

finer

than

2¿¿m

gives

an

indication

of the

plasticity of the purely

clay-sized

portion of the soil and is called the

activity .

Kaolinite has an activity of

0.3-0.5;

1;

ilute of

~0.9;

and

montmorillonite of greater

than 1.5. These valúes

hold

true

not

only

for th e activity of the puré clay minerals but also for coarser-grained

soils whose clay fraction is composed of these minerals. A high

activity is associated w ith those clay minerals tha t can adsorb large

amounts of water within their mineral lattice, and is related to the

chemistry

of the

clay pa rticles. This penetration

of the

clay

m inerals

by wa ter molecules causes an increase in vo lume of the clay minerals,

so that the soil swells. Th us, ac tivity is a mea sure of the prop ensity of a



12 CORRELA TIONS OF SOIL PROPERTIES

clay to

swell

in the

presence

of

water

and may be

used

to i

expansive

clays.

In a less

precise manner, swelling

and

shrinkage

properties

are

also

related to the liquid limit, so that this too can be

used to help identify expansive clays. This is discussed in

Chapter

8.

In broad term s, the plasticity Índ ex reflects the ratio of clay m ineral

to

silt

and fine

sand

in a

soil, tha t

is the

proportion

of

clay m inerals

in

the fines. Since th e silt-, sand- and clay sized particles each nave th eir

characteristic angles

of

internal friction, their relative proportions

largely

determine the angle of internal frict ion,

f )

T

,

and henee to a

large exten t the angle o f

efíective

shearing resistance,

< / > )

o f clay soils.

Thus there are, perhaps surprisingly, correlations of

< p

r and

with

plasticity índex.

These

are

given

in

Chapter

6.

1 2 5 Limitations on the use of consistency limits

It

can be seen íhat,

like

grading, the Atterberg limits are poteníially

related to a w ide

variety

o f soil prop erties.

That

this has been fou nd to

be true , gives ampie just if ícation for the use of grading and plasticity

properties

in the

soil classifícation systems. However,

a l though

Atterberg l imits do enable intriguingly good predictions for some

engineering properties, certain lim itatio ns m ust be recognised. L im it

tests

a re performed on the m aterial fíner than 425jUm, and the degree

to w hich this fractio n reflects the

properties

of the soil will depend on

the

proporíion of coarse material present and on the precise grading

of the soil.

Another l imitat ion

is

that

th e

limit tests

are

performed

on

remoulded soils and the correlations are not generally valid for

undisturbed soils unless the soil

properties

do not change substan-

tially

during remoulding.

This is the

case with

m a n y

nor-

mally-consolidated

clays but the properties of

over-consolidated

clays, sensitive clays and cemented soils

often

differ markedly

from

those predicted from Atterberg limit tests.

«•





14 CORREL ATIONS OF SOIL PROPERTIES

In

t h i s

respect, c lassif ícat ion sys tem s are m ore applicable w here soi ls

are used in remoulded

form

tha n w here they are used in thei r na tura l

s ta te and i t i s not surpr is ing tha t the m o s t comm only used engineer-

ing soil

classifícat ion systems were

all

developed

fo r

e a r t hworks ,

highways

o r

a i rpor ts work.

2.1

COMMON SOIL CLASSIFÍCATION

SYSTEMS

Th e mo st widely used engineer ing so il c lass i f íca t ion systems thro ugh -

out the

English-speaking wor ld

are the Uniííed

system

and the

American

Asso c ia t ion o f S t a te Highw ay and T ranspor ta t ion

Offíciáis

(AASHTO ) system. Of these, the

Unified

system is the mo re generally

applicable

and

more widely used.

I t was

developed f rom

a

system

proposed

by

Casagrande (1948)

and

referred

to as the

Airfield

Classif ícaíion S ystem . Coarse-grained soi ls (sands

and

gravéis)

are

classifíed according to their grading, an d fine-grained soils (silts and

clays)

a nd

organic soi ls

are

classifíed according

to

their plast ici ty,

a s

indicated in

Table

2.1. Classifícat ion i s carried out using particle size

dis t r ibut ion data and l iquid l imit and plast ici ty índex valúes, as

shown in Table 2.2. An ingenious feature of the system is the

differentiat ion

o f

silts

and

clays

by

means

of the

plast ici ty chart ,

included in the

table.

The posi t ion of the A-line was fíxed by

Casagrande, based

on

empirical data.

Th e only

m odif ícat ion f rom

Casa grande s original proposa l is the

smal l

devia t ion a t the lower

end.

The

system

ca n

also

be

used

to

classify soils using only

fíeld

ident i f íca t ion, as

indicated

in

Table 2.3.

An advantage of the system is t h a t i t can be easily extended to

include more soi l groups, giving a fíner degree of classifícat ion i f

required.

The A merican Ass ocia t ion for Test ing and M ater ia ls hav e a dopted

th e

U nified system as a basis for the ASTM soil classifícation, entitled

S tandard Test M eth od for Classifícation o f Soils for Engineering

Purposes , designation D2487. T h e p resentat ion is s omewha t

difíer-

ent

from

t h a t

o f the

U nified system

but the raethod of

classifícatio n

is

almost identical. Th e ma in differences a re t h a t th e ASTM classifíca-

tion D2487 requires classifícation tests to be rformed whereas th e

Unifíed system allows a t enta t ive classifíca; based on visual

inspection only;

and the

ASTM system gives

a

subdivis ión

of the

groups wh ich produces

a

rigidly

specifíed ñ a m e fo r

each

soil

type.

T he

main soi l classifícat ion chart

is

given

in

Table

2 .4 and the

ASTM

versión

of the

soi l plast ici ty ch art

i s

given

in

Figure

2.1.

D efíni tions

of

th e soil descriptions used are given in Table 2.5. The coefíicient of



SOI L CLASSI FI CATI ON SYSTEMS 15

Table 2 1 TH E U N I F I E D S O I L C L A S S I F I C A T I O N S Y S T E M : B A S I C S O I L G R O U P I N G S

Majar divisions

jg

'o

^ S

'3

3

X

.

ftj

^J

3

•*••»

1

^j

Q

^j

Sí

e;

"S

Ijl

líl

djl

o

"a

I

•

Q

C* 3

^ s j E

"^S

'r»

^

1

s '^

a

v^ s:

»££

Su

^~*' 1\

§

§ -s:

'S

^

*

+

C^ s ^f*

^S

C

"S

-2 ^

» ^ j ^

^

V Í3 *

"^ "3

S .§

=3 - J

jf

1

y

i

~SS

J

e

•g

<í o1

1§

0 S

S

w

i

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f

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a

e

a

e

a

m

o

o

f

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¿o ~~

"^ .§ a

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Highly

organic soils

Typical ñames

W e l l

graded gra vé i s , g rave l - s a nd

m i x t u r e s ,

li t t le or no f ines

Poor ly graded g ravé is , g rave l - s a nd

m i x t u r e s , little

or no f ines

Silty

g ravé i s , poor ly graded

gravel-sand-si l t

mix tures

Clayey

g ravé i s , poor ly graded

gravel-sand-clay mix tures

Wel l graded s ands , grave l ly s ands ,

little or no fines

Poor ly graded s ands , gra ve l ly

sands,

little

or no fines

Silty

s ands , poor ly graded

sand-silt mix tures

Clayey

s ands , poor ly graded

sand-clay mix tures

I no rgan i c

silts

and very f ine sands ,

rock f lour ,

silty

or clayey fine

sands with s l ight plas t ic i ty

Inorgan ic c l ays

of low to

médium

plast ici ty ,

gravel ly c lays , sandy

clays, sil ty clays, lean clays

Organic s i l t s

a nd

organ ic

silt-

clays

of low

plas t ic i ty

Inorganic s i l t s , micaceous

or

d i c tomac eous

fine

s andy

or

sil ty

soils , elastic

sil ts

Inorganic c lays of high plas t ic i ty,

fa t

c lays

Organic c lays

of

mé d i um

to

high

plast ici ty

Peat

and

other highly

organic soils

Group

symbols

G W

GP

G M

GC

SW

SP

SM

SC

M L

CL

OL

M H

CH

OH

Pt



U se

grain size curve

in

identifying

th e

fractions

as

given under f ield

Identification

P l a t t l c i t y

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» ~

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r ~ ? r

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O • -•--«••— ' ' 1 ' -

•

s *J

\ ^

m \

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a n d sand

f ro m g r a i n s i ze c u r ve . D epe

percentage of f ines ( f r a c t i o n s m a l l e r than 75/ ím sieve size) c o a r se g r a i ned

classified a s f o l l o w s:

Less than 5

G W , G P ,

SW ,

SP

More t h a n 12% GM, GC, SM, SC

5 to 12%

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o ^ 5' § a

r l

»^

^J •

w

( *•

18

ir 5 fí

^Ü

e n

Sr

3 _

m 3 .

|± -

o

^

" S

^ r "

M ^ .

ftj

*-3

*C

c

w P- 2.

E - S -j

3-

O

n

o

S

S o

^

*-»•

rt i

re

r-f

fc»

•

3

O Q

Si

O Q

i~ i

C L

P

r

o

3

*- t

£

C_

i - t

re

3

3

• .

rt-

o n

i—

» ^

O

O

13

**•

to

1

O

X

0



SOIL

CLASSIFICATION SYSTEMS 17

70

60

I

5

X

40

_ > .

o

30

5

<

a20

10

7

4

FOR CLASSIFICATION OF FINE-GRAINED SOILS AND FINE-GRAINED

FRACTION

OF COARSE-GRAINED

SOILS

Equation

of

.Horizontal

then PI=O.7

.Equation

oí

Vertical

at

then Pl=0.í

A--IÍ

at

Pl='

3(LL-

IT-I

LL=16

KLL-fi

CL-ML

I

ne

\

2O)

ne

to Pl

=

)

¿

í

ox

^

Lo

-25.5

y

ov,

rOL

1 j

\°*

vJ/>

^

z

,

*

v

>

s

J

X

MH

°

.

»v

0

/

or

s

OH

y

1O 2 3 4 S 6 7 8 9 1 O 11 12

Liquidlimit LL)

Figure 2.1 Soil plasticity chart used with the STM an d

Unified

soil classifi-

cation sysíems

uni formi ty ,

Cu and the

coefficient

of cu rva tu r e ,

Ce,

of the grad ing

curve ,

wh i ch

are

used

in the

classif ication,

are

defined

in

Table 2.4.

The soil

ñames used

for

each

of the

soil group s

are

defíned

in

Tables

2.6,

2.7 and

2.8.

The Bri t ish Standard classification

system

BS 5930) is , like th e

Unified system, also based

on the

Casagrande classification

but the

definitions

of sand and gravel are slightly different, to be in keeping

with

o the r

Brit ish Standards, and the f ine-grained

soils

are divided

into fíve plastici ty ranges rather than the simple

low

and high

divisions of the Unified and the original Casagrande systems. In

addi t ion, a considerable

n u m b e r

of sub-groups have been i n troduced .

The basic soil ñam es, symbols and qualifying t e rms are given in Table

2.9 and the definit ions of the soil groups and sub-groups can be

obtained

from

Table

2.10

in

conjunct ion with

the BS

versión

of the

plastici ty chart , Figure 2.2.

It can be seen that both the ASTM and, part icularly, the BS soil

classification systems subdivide the soil in to a m uc h larger num ber of

group s th an the earl ier system s. Al tho ugh

this

allows a m ore precise

classification,

it

negates

two of the

main attr ibutes

of the

Unified

classificat ion: the

system s

are no t

long er simple

and

easy

to

remem ber

but require constant reference to a table and ch a r t; and they canno t

be implemented without recourse to laboratory test ing.




ble 2 .3 TH E

U N I F I E D S O I L C L A S S I F I C A T I O N S Y S T E M : F I E L D I D E N T I F I C A T I O N

Field identiflcation procedures

(Excluding par ti

cíes

larger than 75mm and basing fractions on

estimated weights)

t

1

i~

-Ü *

^

'3

.2

¿

í I 'S

lis

i

2^5= °

?í|l

2 J a

8

•* =

a e

3

J**

Ib

.O

o

:s

ís

o

o

Ui

«I

O.

W )

ü

13

o

u

.n

3

O

ja

«

^ .< £

1 .s

1

X

«,

S

V,

— .N U

Í3

O

o;

'55

2 £ S E

« o -S i

= * •

. 1

«

£

¡>t

j j

< * x ¡ o f _

ia C

;

S -s

c

^

§

•«

S u

f

^ < ¿

£ a ^3

, I

•- _ ^ . s ^

Q1

^ 5j

-C

1%I«

s -

2 • * = < - . £

s

J a

£ .u -S

-5 .0 - s; .

H f l

5

•*

S

M

o

h

h

o

c

f

r

a

o

s

s

m

e

h

4

7

m

m

s

e

o

r

v

s

u

c

a

c

o

h

e

v

e

o

e»

1§

|o?

§ ^ ^ .

•5; :

{ j -^

-c: ^-

= - - ' < = •

n <ú í- 5C aj

1*11*

3

-a

o

JU

^ 0 3

a

« j

,a

g^^

J

*-.

o -S

-«

-2:^

.•S -c o ^ ^

- ^ . .a ^ -

-^

¿

^

c 2

1 4 I H

o _g Q

W i d e r an g e i n g r a i n s i z e an d s u b s t an t i a l

a m o u n t s o f a ll

i n t e r m e d í a t e

particle sizes

Pred om ina ntly one s ize or a range of s izes

w i t h

s o m e i n t e r m e d í a t e s i z e s m i s s i n g

N o n - p l a s t i c fines ( f o r

id e n t i f lc a t io n

p r o c e d u r e s ,

s ee ML b e l o w )

Plás t ic f ines (for id e n t i f ic a t io n p r o c e d u r e s , s e e C L

b e l o w )

W i d e r an g e in grain sizes a n d s u b s t a n t i a l

a m o u n t s o f a ll

i n t e r m e d i ó t e pa r í ic le

sizes

P r e d o m i n a n t l y o ne size o r a range o f s i z es w i t h

s o m e i n t e r m e d í a t e

size

m i s s i n g

N o n - p l a s t i c

f i nes

( f o r

id e n t i f ic a t io n

p r o c e d u r e s ,

s ee

M L

b e l o w )

Plás t ic f ines (for id e n t i f ic a t io n procedures , see CL

b e l o w )

Identification procedures on fraction smaller íhan 425um sieve

h's;

- ¿ l l

§:s-s

a.|.g

§ ~ "

^

o

E * -

0

3

-.3 -.

G .g J

li|

Í3

Q

:

« u

o,

Highly

organic soils

Dry sírength

(crushing

charac-

teristics)

None

to

s l ight

M é d i u m

to

high

Slight to

m é d i u m

Slight to

m é d i u m

H i g h to

very high

M é d i u m

to

high

Dilatancy

(reaction

to shaking)

Q u i c k to

s low

None to

very

s l o w

Slow

Slow to

n o n e

None

N o n e

to

very slow

Toughness

(consistency

near plástic

limit)

None

M é d i u m

Slight

Sl ight to

m é d i u m

H i g h

Slight

to

•sdium

R e ad i l y

identif ied

b y c o l o u r, odou; p o n g y

feel

a n d

f r e q u e n t l y

b y f i b r o u s

t e x t u r e

Group

symbols

G W

G P

G M

G C

SW

S P

S M

SC

M L

C L

O L

M H

CH

O H

Pt






ble 2.5

SYSTEM

DEFINITIONS OF SOIL DE SCRIPTIONS FOR THE ASTM SOIL CLASSIFICATION

Description

Defm ition of

material

Boulders Reta ined on

300mm

(12in)

sieve

Cobbles Passing 300mm (12in); retained on 75mm (Sin) sieves

Gravel Passing 75mm (Sin): reíained on 4.75mm (No. 4) sieves

coarse Passing 75mm (Sin); retained on 19mm

(|in)

sieves

fine Passing 19mm

(|in);

retained on 4.75mm (No. 4) sieves

Sand Passing 4.75mm (No. 4); retained on 75/zm (No. 200) sieves

coarse Passing 4.75mm (No.

4);

retained

on 2mm

(No.

10)

sieves

médium

Passing

2m m (No. 10); retained on 425/mi (No. 40) sieves

fine

Passing

425/¿m

(No. 40); retained

on 75/¿m

(No. 200) sieves

Clay

Passing 75/mi

(No. 200) sieve that

can be

made

to

exhibit plasíicity

within

a range of water contents and that , exh ibits considerable

s t rength when

air

dry .

F or

classification,

a

clay

is a

fine-grained

soil,

or fine-grained

port ion

of a

soil, w ith

a

plasticity

índex

of

equal

to or

greaíer than 4, and ploís above íhe

A line

on íhe plasíicity charí.

Silt Passing 75/^m (No. 200) that

is

nonplastic

or

ve ry slightly plástic

and

exhibits little or no dry strength when a ir dry. For classification, silt

is

a fine-grained

soil,

or fine-grained

portion

of a

soil, w ith

a

plasticity

Índex

less than 4 or which ploís below th e A line on the plasticity

charí .

Organic

clay A clay or sill with sufficient organic conlent to

influence

íhe soil

or sill properlies. For classification, an organic clay or silt is a soil that

would be classified as a

clay

or silí

excepl

Ihat ils liquid limil

valué

afler oven drying

is

less Ihan

75 of ils

liquid limií before oven

drying.

Peat

A soil

composed

of vegetable tissue in

various

stages

of

decomposi-

lion usual ly w ilh an organic odour , a dark-brow n lo black colour, a

spongy consislency

and a

lexture ranging

from fibrous lo

amor-

phous .

*

Sieve sizes

and numbers refer to

U.S.

square

sieves.

As a

result

of the introduction of

these classification systems,

a

subtle change

has

arisen

in the defmition of

silt. Normally, silt

and

clay particles are

defíned

by their particle size, the división betw een

silt

and clay being 5/rni in the ASTM and AASHTO

defmitions,

and

2/mi

in the BS

defmition.

The

plasticity chart

was a

useful

way of

separating silts from clays, which worked for mosí soils: clays

generally plotted above

the

A-line

and

silts

below üthough

excep-

tional clays were known to plot below it. Now, ,

r

classification

purposes,

whether

a

soil

is a silt or a

clay

is defíned in terms of

w hether

it plots above

or

below

the

A -line, rathe r than

o n its

particle sie.

T he

British Standard system suggests that, to avoid confusión, the term

M-soil

is used for those fine-grained soils that plot below the A-line,

but this does not

seem

to ha ve gained popular acceptance.



SOIL CLASSIFICATION SYSTEMS 21

70

6

= 4

2

er

3 3

20

10

SILT M-SOIL), M, plots bolowA-line\y becombinedas

CLAY C, plots above

A-line

/ FINE SOIL, F.

L - Low

plasticity

U

- Uppor

plasticity rango

i

Inter-

medíate

NOTE: the

letter

O is added

to the symboi of any

material

-

containing

a

significant

proportion

of

organic

mat ter

e.g. MHO

CL

ML

x

Cl

MI

H -

High

CH X

-

M«J

m

V -

Very

high

C

:V

MV

E - Extremely

high

x

°>

ME

O 1 2 3 4O 5 6 7 8 9O 1 11 12

Plasticity indox ( )

Figure

2 2

oilplasiicity

chart used with the British Standa rd so il

classification

system

Although the

Casagrande- type systems c lassify

soils

aceo r d ing

to

the i r e ng inee r ing prope r t i e s ,

t h ey

are no t s t r ic t ly inte r pre t ive , in that

they

do not over t ly

classify soils

as

good

or

bad

for a

part icular use

How eve r , they can be re adi ly used in th is way with the a id of tables or

charts such

as

those indicated

in

Tables 2 11

and

2 12

The AASHTO soi l classif ication system M 145) doe s not classify

soils by type i .e . sands, clays etc .) but sim ply divides them into seve n

major groups, as shown in

Table

2.13. Groups A-1, A -2 and A-7 are

usually

subdivided as indicated.

Typical

mate r ia ls in

each

g ro up a r e

indicated in Table 2 14 Although soils are divided into granular

mate r ia ls groups A-1, A-2 and A-3) and

silt-clay

mate r ia ls groups

A-4

to A-7), th e dist inct ion is

less

c lear-cut than with the Casag-

rande-type systems. This i s part icularly t rue of the A-2 group, which

c an

include soils with

a

considerable sil t

or clay

content . Clays

are

dist ingushed f rom silts

on the

basis tha t c lays have

a

plasicity Índe x

of

greate r than 10: unlike th e A-line división of the Casagrande

plasticity

char t , th is ra the r arbi t rary divis ión does

no

truly dist in-

guish between

these tw o

types

of

soils.

Also,

organic soils

are not

included in the c lass i f icat ion. However , the system must be judged

aceording to its own aims, which ar e

specifically

to assess th e



b le 2 6

F L O W C H A R T

F O R

C L A S S I F Y I N G C O A R S E - G R A I N E D S O I L S ( M O R E T H A N

s o

R E T A I N E D

O N

is^m S I E V

<5 fines

and

an d / o r l>Cc>3

G R A V E L

% grave l>

%san d

^5-12% fines

and

Cu<4

and/or l>Cc>3

12 %

fines

fines-ML

o r MH

fines-CL, CH,—

(o r C L - M L )

f ínes-ML or M H

f ines-CL

or C H ,

(or C L - M L )

f ines-ML

o r MH

fines CL or CH

f ínes-CL-ML

> G W - G M

> G W - G C

+ G P - G M

> G P - G C

*G M

< 15% s a n d -

S í l 5 % s a n d -

< 1 5% s a n d -

>

15%

s a n d -

<

15% sand

-

5=15% sand

- » < 1 5 % s a n d -

> 1 5 % s a n d -

< 1 5 % s a n d

> 1 5 % s a n d -

< 15%

sand

*G C

> G C - G M

15%

sand

-

^

15%

s a n d -

->

< 15% s a n d -

^

>

15% s a nd -

< 1 5%

s a n d -

' 5= 15% s a nd -

>W

'We

>Po

> P o

>W

>W

>W

>W

>Po

>Po

>Po

»Po

Si

» S i l

••Cl

>Cl

v S i l

> S i l



•• • • •

n n n

• •

B

1 1 E 1 1 1 1 I I I» »

SAND

,<5 fines

5-12

fines

Cu

6 and 1

< Ce <

3

Cu<6 and/or 1

>Cc>3

Cu^óand l<Cc<3

, T

J

* fines-ML

or MH —

X

ines-CL, CH,

(o r CL-ML)

— ovv

- ^

> S T p

—

+SW-SM

-:

—

»sw-sc

'-^- 1J 70 giavci

~^

15 gravel

l jf*1 1 AOTIVPl

•> 15

grave l

—

—

> <

15

gravel

^ 15%

gravel

:—

-» < 1 5 gravel

>15 gravel

Cu<6

and/or

l>Cc>3x,

>12

fines

,

fines-ML or MH >SP-SM

fines-CL or

CH >SP-SC

(or CL-ML)

fines-ML or

MH -S M

fines-CL-CH

> SC

fines L ML

->SC-SM

15 g r a v e l -

' ^15 g r a v e l -

» <

15

g r a v e l -

' ^ 1 5 grave l -

< 15

grave l

15

grave l

< 15 gr ave l

<15 g rav e l -

15

grave l

•

W

W

» P

>

»W

*

»W

»W

*P

••

••

- >

- >

-*

->

-*

->



Table 2 . 7

F L O W

C H A R T F O R C L A S S I F Y I N G I N O R G A N I C F I N E - G R A I N E D

S O I L S

( 5 0 % O R M O R E P A S S E S

7 5 / « n

S I E V

GROUP SYMBOL

I n o r g a n i c

L L < 5 0

P I > 7 a n d

plo t s o n o r above

'A ' - l ine

4<PI<7

a n d

>C L - M I

p lo t s on or a b o v e

'A ' - l ine

PI<4

o r

p l o t s -

be low 'A ' - l ine

/ LL- ove rd r i e d

Orgahic — . ,<0.75

1 LL-not

d ne d

<3Q

plus

N o . 200^< 1 5

p lus

N o. 200

\5-29 p lus N o . 2C

sand grave l

15-29 p lus N o . 2 0 0 - x — > % sand >

sand

<

p lus No . 200<f

sand

< g r a v e l «

< 1 5

grav

5*15 grav

< 1 5 s a n

l^\5 san

,<30% p l u s

N o .

200<-»<15 p l u s

N o . 200-

sand

N. t

grave l

•

'15-29

p l u s

No. 2(Kk^»

sand

^

/ o

sand <

<15 grav

p lus

No. 200<(

* ^

1 5

g r a v

sand < g r a ve l v^ <

15

san

15 sand

,<30% plus No . 200^-* < 15 p lus No . 200-

15-29% plus No. 200 sand

>

sand <

sand ^% g r á v e l a —

15

g r a

^ 15

g r a

s a n d

<% g r a v e l ^ — + < 15

san

^ 15 s a n

> S e e T a b l e 2 . 8



vi t i i i i i i i i »

M

I J I » » » V V I f t f t i i

ft}11Il

Inorganic

PI plots on

or >C H

above

'A'-line

PI

plots below

> M H

'A'-line

/LL-overdried

Organic —

—-j<0.75

—»OH

1

LL-not

dned

<30 plus No.

200^-»<15

plus No. 200

N

5-29

plus

No. 2(XK^

sand

sand

<

30

plus

No

,

sand

gravel

N ,

sand

<

gravel

< 15 grav

\5

grav

< 15 sand

^

15 san

,<30

plus No. 200 -> < 15 plus No.

200

15-29

plus

No. 2(Xk-*

sand

^

sand <

sand < gravel -^— >< 15 gra

15

grav

í

30

plus

N o.

sand < gravel

:15

sand

15

sand

> S e e T a b l e 2.8



ble

2 . 8

F L O W C H A R T

F O R

C L A S S I F Y I N G O R G A N I C F I N E - G R A I N E D S O I L S (5 0

O R

M O R E P A S S E S 7 5 /i m S I E V E )

GROUP

SYMBOL

and

plots on

or

above

'A' - l ine

PI<4 or plots

below 'A'-line

OH

Plots on or

above 'A'-line

Plots below,

'A ' - l ine

<30 plus

No. 200-

5 = 3 0 %

plus No. 200

<30 plus

No.

200

S s 30% plus No. 2

<30

plus

No.

200-

Ss 30 plus No. 200

,<30% plus No. 200-

•> 30 plus No. 2

<15 plus No. 200-

15-29 plus No.

200-=

sand

>

graveé

sand < g r a v e l -

> < 1 5 % plus No. 200

15-29 plus No. 20(k

sand

^

gravel -

sand

< gravel

> < 1 5 % p lus No. 200

'15-29

plus

No. 200-

sand ^

grave l

•

sand

<

grave l .

•<15 plus No. 200-

'15-29 plus No. 200-

sand

>

grave l -

sand

<

gravel

sand

gravel

'

sand < grave l

< 15 gravel

5* 15% g rave l

-»<15 sand

• > 1 5 %

sand

sand

¿t

gravel

% sand < % gravel

< 15 gravel

>15 g r a v e l —

<15

sand

Sil5

sand

sand

5s

gravel

sand < gravel

K15

gravel

•>15 gravel

-*<15

sand

1 5 %

s a n d —

-» sand

^

gravel

sand < gravel

<

15

gravel

15 %

gravel —

<15

sand



SOIL CLASSIFICATION SYSTEMS 27

Table

2 9

Ñ M E S AND D E S C R I P T IV E L E T T E R S FOR G R D I N G AND P L S T I C I T Y

H R TERISTI S

escriptive ñame Letter

Main íerms

Qualifying terms

Main terms

Qualifying terms

Main term

Qual i fy ing

term

G R A V E L

SAND

Well graded

Poorly

graded

Uniform

Gap

graded

FINE SOIL, FINES

m ay

be differentiated in to M or C

SILT (M-SOIL)

plots below A-line

of plasticity chart

(of restricted

plástic

range)

CLAY

plots abo ve A-line

(fully

plástic)

Of

lo w plasticity

Of

intermedíate

plasíicity

Of high plasticity

Of very high plastisity

Of extremely

h igh

plasticity

Of

upper

plasticity

range*

incorporating groups

I, H, V and E

PEAT

Organic

may be suffixed to any

g roup

G

S

W

P

Pu

Pg

F

M

C

L

I

H

V

E

U

Pt

O

This term

is a

useful guide when

it is not

possible

or not required to desígnate the range of

l iquid l imit more closely,

e.g. during the rapid

assessment

of soils.

suitabili ty of

soils

fo r

pavement

subgrades; the

higher group num bers

being progressively less suitable.

In

this way the system is more

restricted yet more interpretive than the Casagrande-type systems,

since

it not

only

classifíes

soils into groups

o f

similar properties

but

also passes ju dgem ent abou t the qua lity or suitab ility of the soils in

each group.

A

further

refínement of the

AASHTO system

in

this

respect is the use of a

group

Índex , to ev alúa te subgrade quality . It is

calculated from

the

formula :

Group

index =

( JF-35)[0.2 + 0.005(LL-40)]-

r

-0.01(F-15)(P/-10)

where

F is the

percentage passing 0.075mm sieve, expressed

as a

and

whole

number .

This

percentage

material passing

th e

75mm sieve

LL

is the

liqu id limit

PI is the plasticity Índex.

is based only

on the



oarse

soils

(<35% fines)

Sands (>50% of coarse

material is of

sand size

-

<

2mm)

— <

P E?. 2- E £

1

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Table

2 11

E N G I N E E R I N G P R O P E R T IE S

O F

C O M P A C T E D

S O I L S ,

C L A S S IF IE D A C C O R D I N G

T O T H E

U N I F I E D

S Y S T E M

Ty ÍCal

ames

ofsoil

groups

Important

engineering properties

Relative desirability

No.

1 is

consid

Rolled

Earthfill

dams

Canal

seclions

Group

symbols

Shear

Permeabili ty strenglh ibility

when

compacted when

compacted

an d

compacted

salurated

saturated

Workability Homo-

as a qeneous

,

conslruclion embank-

material mení

Core Shell

Com-

Erosión S

pacled .

resist-

i

earth

anee

l

lining

Well-graded gravéis,

g rave l -sand mix tures ,

little

or no fines

GW

Poorly graded gravéis,

GP

gravel-sand mix tures ,

l i t t le

or no fines

Sil ty

gravéis , poorly G M

graded gravel -sand-si l t

m i x t u r e s

Clayey gravéis ,

poorly G C

graded grave l -sand-c l ay

mixtures

Wel l -graded

sands

S W

gravel ly

sands

l i t t le

c1

no f ines

Poorly g raded sands , S P

gravelly sands

l i t t le or

no f ines.

Perv ious Exce l l en t Neg l ig ib l e Exce l l en t —

Very Good

pervious

Semiperv ious

Good

to i rnperv ious

I m p e r v i o u s Good

to

fair

Negl igible Good

Negl igible Good

V ery lo w Good

2

Perv ious Exce l en t Negl ig ib l e Exce l l en t — —

Pervious Good Very lo w

Fair

3

If

6

gravel ly

4 7

I f

If

gravelly g rave l ly

2



SOIL

CLASSIFICATION SYSTEMS 31

ÍN r^

2

o

• < ñ

o

s

s

• W

o

O 3

u

o

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r

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c

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E

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S s

8

3

2

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e c

E

0

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a

o

3 3

cS

2

VI

3

.2

B

u

ai

C

l

a

s

p

y

g

a

s

n

c

a

m

i

x

u

e

i*

3

U

E

3

•3

4>

S

1

3 3

o .2

> £ ;

II

Ji

S

j

S

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s

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r t _ ^ j

a

g

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w

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c u 5

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g o.

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UH

2

3

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E

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n

g

n

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c

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m

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d

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r

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n

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g

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P

a

n

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h

h

g

y

o

g

n

c

s

o

s



Table 2.12 E N G I N E E R I N G P R O P E R T IE S O F C O M P A C T E D S O I L S , C L A S S I F IE D A C C O R D I N G T O T H E E X T E N D E D C A

A F T E R CP2001: B S I 1957

Casagrande

group-

symbol

G W

GC

G U

GP

GF

S W

S C

S U

SP

Valué

as a

road

foundation when

not subject tofrost

action

E x c e l l e n t

Exce l l en t

Good

Good

to

excellent

Good to excel lent

E x c e l l e n t

to

good

E x c e l l e n t to good

Fair

Fair to

good

Potential frost

action

Non to v ery s l ight

M é d i u m

N o n e

on to very

slight

S l ight

to

médium

N o n e

to

v e r y

_t

•

i

4.

c

i gh t

'- 'o

M é d i u m

N o n e to

very

i • i

slight

None

to

very

slight

Shrinkage or

swelling

properties

Almos t none

V e r y s l ight

Almos t none

Almos t n o n e

A l m os t n on e

to slight

A l m os t n on e

V e r y s l igh t

A l m os t n on e

Almost n o n e

Drainage

characteristics

Excel l en t

Prac t i ca l ly

i m p e r v i ou s

E x c e l l e n t

Excel lent

F a i r

to

pract ical ly

i m p e r v i ou s

Exce l l en t

P rac t i ca l ly

i m p e r v i o u s

Exce l l en t

Excel lent

Bulk

dry

at opt imu

compactio

Ib .

/cu .

f t

voids

rati

> 1 2 5

< ? < 0 . 3 5

> 1 3 0

e<0 30

e<0 50

e<0 45

> 1 2 0

e < 0.40

> 1 2 0

f\

r\ <Ü

>125

e < 0 . 3 5

>100

0.70

>100

e

< 0.70



SF

M L

CL

OL

M I

CI

Oí

M H

CH

OH

Pt

Note .

L -

I-

H

-

Fair

to good

Fair to poor

Fair

to poor

Poor

Fair to

poor

Fair to

poor

Poor

Poor

Poor to very póor

Very poor

Extremely

poor

Group symbols as for Unif ied system

low plastici ty ,

P I

less than

35

intermedíate plastici ty , PI

35-50%

high plastici ty ,

P I

greater than

50

Slight to high

Médium to very

high

Médium to high

Médium to high

Médium

Slight

Slight

Médium to high

Very slight

Very slight

Slight

except

fo r

plactici ty ranges:

b

Almost n o n e

to médium

Slight

to

médium

Médium

Médium

to

•

v

high

Médium to

1 1

high

High

High

High

High

High

Very

high

Fair

to practically

impervious

Fair to poor

Practically

impervious

Poor

Fair to

poor

Fair to practically

impervious

Fair to practically

impervious

Poor

Practically

imperv ious

Practially

impervious

Fair to poor

-

> 105

e

0.60

i o o

e<0.70

i o o

e<0.70

>90

e 0.90

i o o

e 0.70

>95

e

0.80

>95

e

0.80

> 1 00

>90

e < 0 . 9 0

i o o

e

0.70

—



• • . . . .

Table

2.13

A A S H T O S O I L C L A S S I F IC A T I O N S Y S T E M M

145)

,-, , , . ~

Granular materials

General

classmcatwn

,->cn/

• T C \

or

less passing /jum

4-7 4-3 4-2

Group classification

A l a A l b 4-2-4 4-2-5 4-2-6 4-2-7

Sieve analysis:

Percentage passing:

2 m m 5 0 m a x — — — — — —

425/rni 30 max 50 max 51 min — — — —

75/¿m 15 max 25 max 10 max 35 max 35 max 35 max 35 max

Charater ist ics of

f rac t ion passing

425/im:

Liquid

l imit — — 40 max 41 min 40 max 41 min

Plast ici ty índex 6 max NP 10 max 10 max

11

min

11

min

Group índex

-

typical valúes 0 0

0 4 max

Usual

types of

Stone

f r agment s Fine Silty or clayey

gravel

and sarid

significant

gravel

and

sand

s a n d -

c on s t i t u e n t materials

. • .

.

Genera l r a t ing as

subgrade Excel l en t

to

good

Si

(More

A-4

36 min

40 max

10

max

8 max

Silty

Fa

* Plast ici ty

Índex

of A -7 -5 s u b g r ou p is e q u a l to or less t h a n LL m i n u s 30 . Plast ici ty índex of A -7 -6 s u b g r ou p i s g rea te r t h a n LL m i n u s 30 .



SOIL

CLASSIFICATION

SYSTEMS 35

Table 2 14

D E S C R I P T I O N S O F S O I L T Y P E S I N T H E S H T O S O I L C L S S I F I C T I O N

S Y S T E M

Classification

of

materials

in the

v arious groups applies

only to the

fract ion passing

th e

75mm sieve. The proportions of boulder and cobb le-sized particles should be recorded

separately

and any specification

regarding

the use of A - l , A-2 or

A-3 materials

in

construction

should

state

whether

boulders are

permitted.

ranular

materials

Silty clay

materials

Group A-l. Typical ly

a well

graded

mixture o f

stone fragments

or

gravel, coarse to fine sand and a

nonplas t ic or

feebly

plástic

soil

binder .

However ,

this g roup

also

includes stone fragments , gravel ,

coarse

sand, volcanic cinders, etc.

wi t hou t

soil binder.

Su b g rou p

A-l-a is

p re d ominan t l y

stone

fragments

o r

gravel ,

with

or

w i t h o u t

binder.

Subgroup A-l -b

is

predominant ly

coarse sand with or w i tho ut b inder .

Group

A-3. Ty pically fine beach sand

or

desert sand without

silty or

clayey fines or with a very small

proport ion of nonplast ic s i l t . The

group also includes stream-deposi-

ted mixtures of poorly graded fine

sand with limi ted amoun ts

of

coarse

sand

and gravel.

Group

A-2. Includes a w ide varie ty of

granular materials which are bor-

derline between the granular A-l

and A-3 groups and the silty-clay

materials

of

groups

A-4 to

A-7.

It

includes

al l

m aterials with

not

more

than

35 fines

which

are too

plás-

tic or have too m a n y fines to be

classified as A-l or

A-3.

Subgroups A-2-4 and A-2-5 include

various granular materials whose

finer

particles (0.425mm down)

have íhe characteristics

of the A-4

and A-5

groups , respectively.

Subgroups A-2-6 and

A-2-7

are simi-

lar to

those

described above but

who se finer particles have the ch ar-

acteristics of A-6 and A-7

groups,

respectively.

G rou p A-4. Typically a nonplast ic or

modera te ly plástic silty soil usual ly

with a high

percentage passing

th e

0.075mm sieve. The group also in-

cludes

mixtures

of

silty

fine

sands

and silty gravelly sands.

Group A-5. Similar to material de-

scribed

under group

A -4

except tha t

it is usually diatomaceous or

micaceous

and may be

elastic

as

indicated

by the

high l iquid

limit.

Gr ou p A-6. Typical ly

a

plástic clay

soil having a high percentage pas-

sing th e

0.075mm sieve. Also m ix-

tures of clayey soil with sand and

fine gravel .

Materials

in

this group

have a high volume change between

wet

and dry states.

Group A-7. Similar

to

material

de-

scribed

under

group

A-6

except tha t

it has the

high liquid limit

charac-

teristic

of group A-5 and may be

elastic

as

well

as

subject

ío

high

volume change.

Subgroup A-7-5 materials have m od-

érate plasticity Índices in relation to

the liquid limits and may be

highly

elastic

as

well

as

subject

to

volume

change.

Subgroup

A-7-6 materials

have high

plasticity

Índices in relation to the

liquid limits

and are

subject

to

extremely high volume change.

Gro up A-8. Includes highly organ ic

materials. Classification

of these

materials is based on visual inspec-

tion

and is not

related

to grading or

plasticity.



36

C O R R E L A T I O N S

O F

SOIL PROPERTIES

Table 2 15 C O M P R I S O N

O F

S O IL G R O U P

I N

U N I F I E D S Y S T E M

Group

G-F

GF

S

S-F

SF

FG

FS

F

Pt

S system

Subgroup

Subdivisión

GW

GP

G-M

G-C

GM

GC

sw

SP

S-M

S-C

SM

SC

MG

CG

MS

CS

M

C

GPu

GPg

GW M

GPM

GWC

GPC

SP u

SPg

SW M

SP M

SW C

SPC

MLG,

MIG

MHG, MVG,

M EG

CLG, CIG

CHG, CVG,

CEG

MLS, MIS,

MHS, MVS,

M ES

CLS,

CIS

CHS, CVS, CES

M L, M I

M H ,

MV, ME

CL,

CI

CH, CV, CE

Comparable soil group

in n i f i e d system

Most probable

Possible

GW

GP

GP

GW-GM

GP-GM

GW-GC

GP-GC

GM

GC

SW

SP

SP

SW-SM

SP-SM

SW-SC

SP-SC

SM

SC

ML, OL (3 )

M H ,

OH

(3 >

CL'

4

'

CH

(4)

M L, OL

(3)

M H , OH'

3

'

CL (4 >

CH'4'

ML, OL

(3 )

M H , O H (3 )

CL'4'

CH'

4

'

Pt

SW'2'

Sp 2

GW'1'

SP'

2) SW'1 2'

SW-SM'2'

GW-GM'1', SP-SM'2',

SW-SM'

1

2

'

SW-SC'2'

GW-GC'1',

SP-SC'

2

',

SW-SC'1 2'

SM'2'

SC'

2

'

SW'1'

SW-SM'1'

SW-SC'1'

GM<

2)

, SM'2 5'

GC'2', SC'2 5'

SM'5'

SC'5'

-

Notes:

(1 ) These p ossibilities

arise

because soil that is udged to be gap-graded using the BS system may

satisfy

the criterion

Cc=(D30):z/(D10xí)60) =

between

1 and 3 used in the Unified

system.

(2) These possibilities

arise

because

of diflerences in the

definitions

of

sand

and gravel

sizes

between the BS and

Unified

systems.

(3)

Soil

will

be classified into these

groups

if the BS symbol is

suffíxed

with the letter

'O'.

(4 ) Soil will b e

classified into these

groups if it

plots above

the A

line, even

if the BS

symbol

is suffixed

with

the

letter

'O'. However, this will rarely happen.

(5) These possibilities arise because fine soiis are defined as havin g at least 50 fines (< 4 25¿ im) in the Unified

system

bu t

having

at least 35 fines in the BS

system.



SOIL CLASSIFICAT ION SYSTEMS 37

Table 2 16 C O M P A R I S O N

O F

S O I L

G R O U P I N

A A S H T O

S Y S T E M

Soil

group

in

Umfied ASTM

systems

GW

GP

Most

probable

A - l - a

A - l - a

Comparable

soil groups

in SHTO system

Possible

A - l - b

Possible

but

improbable

A-2-4, A-2-5,

A-2-6,

A-2-7

A - 3 A-2-4,

A-2-5, A-2-6,

A - 2 - 7

GM

GC

SW

SP

SM

se

M L

CL

OL

MH

CH

OH

Pt

A-l-b, A-2-4,

A-2-5,

A-2-7

A-2-6,

A-2-7

A- l -b

A - 3

A - l - b

A-l-b, A-2-4,

A-2-5,

A-2-7

A-2-6, A-2-7

A - 4

A-5

A - 6

A-7-6

A - 4

A -5

A-7-5, A-5

A - 7 - 6

A-7-5, A-5

—

A - 2 - 6

A-2-4, A-6

A - l - a

A-l -a

A-2-6,

A-4 ,

A -5

A-2-4, A-6,

A-4, A-7-6

A-6, A-7-5,

A -4

A-6, A-7-5,

A - 7 - 6

—

A - 7 - 5

—

—

A-4, A -5, A-6,

A-7-5, A-7-6,

A- l - a

A - 4

A-7-6,

A - 7 - 5

A-3, A-2-4,

A-2-5, A-2-6,

A - 2 - 7

A-2-4,

A-2-5,

A-2-6, A-2-7

A - 6

A-7-5,

A-7-6, A- l - a

A-7-5

—

—

—

A-7-6

—

A-7-6

—

When applying the formula, the following rules are used:

1 )

When th e calculated group Índex is negative, it is reported as

zero.

2) It is reported to the nearest

whole

number .

3)

When calculating the group

índex

of subgroups A-2-6 and

A-2-7, only

the

plasticity índex port ion

of the

formula should

be used.

The g roup Índex is usually show n in brackets after the group symbol.

Because of the criteria

that

define subgroups A-l-a , A-l-b , A-2-4,

A-2-5

and

group

A 3,

their group índex will always

be

zero,

so the

group índex

is

usually omitted from

th e classification.

Originally the group

índex

was used directly to

obtain

pavement

thickness designs, using th e

group

índex method but this approach



38

CO RRE L A T I O N S OF SOIL PROPERTIES

Table

2 1 7 C O M P A R I S O N OF S O I L GROUPS F R O M TH E A A S H T O TO THE U N I F I E D S Y S T E M S

Soil group

in

S TO

system

A l a

A l b

A-3

A-2-4

A-2-5

A-2-6

A 2 7

A-4

A-5

A-6

A-7-5

A-7-6

Comparable soil

groups

in

Unified ASTM systems

Mos t

probable

G W , G P

S W , S P , G M , S M

SP

G M, SM

G M , SM

GC, SC

G M , G C , S M , S C

ML, OL

O H , M H , M L,

OL

CL

O H , M H

CH , CL

Possible

SW , SP

GP

—

G C, SC

—

G M , S M

—

CL, SM, SC

—

ML, OL, SC

M L , O L , C H

ML, OL, SC

Possible but

improbable

G M , S M

—

SW ,

G P

G W ,

G P, SW, SP

G W , G P , S W , S P

G W ,

G P, SW, SP

G W , G P , S W , S P

G M ,

G C

SM ,

G M

G C, G M , SM

G M , S M , G C , S C

O H ,

M H , G C ,

G M , SM

has now been

superseded

and group índex valúes are used

only

as a

guide.

Numerous other methods of classification have been proposed .

Classifícations aimed specifically at

identifying

expansivo soils and

frost

susceptible soils are

given

in Chapters 8 and 9.

2.2 CORRELATION OF THE

UNIFIED

BS AND AASHTO

SYSTEMS

A correlation between

the BS and

U nified/ASTM systems

is

given

in

Table

2.15. Because

the two

systems

share a common

origin,

it is

possible to correlate the soil groups with a reasonable degree of

confidence. H owever, minor

differences

beíween the systems mean

that the possibility of ambiguity can arise, as explained in th e

accompanying

notes. The totally

different

basis of the A A S H T O

system means that there

is no

direct equivalence between

it and the

groups of the

U nified

system . This is indi cate d in Tables 2.16 and 2.17

which show correlations betw een the U nifíed and AA SH TO systems.

A full comparison of the U nified, AASHTO and now-superseded U S

Federal A viation Agency FA A ) system s is given by Liu 1970). The

FAA soil classification system

is ,

like

the

AASH TO sys tem,

an

interpretive

one in

that soil

is

divided into

a

number

of

classes

according to their suitability as runway subgrades. H owever , the

FA A now uses the U nified system.



Chapter 3

NSITY

3 1

NAT URA L DENSITY

There are two measures of soil density; bulk density which mcludes

th e mass

of

both soil

and

pore water

and dry

density which ignores

th e

efíect

of the

contained water.

The

relationship betw een bu lk

and

dry

densities

is:

where p¿ is the dry density

p

is the

bulk densiíy

and

wn

is the moisture contení .

Bulk density

is

usually

of

prima ry consideration wh ere density

valúes are used directly; to calcúlate earth pressures b ehin d retainin g

walls

or

basements

fo r

exam ple since

it is the

com bined mass

of

soil

and

water that determines th e pressure.

Probably

a

more common

u se of

density

is as a

measure

of the

state

of

packing

of

soil particles and

fo r this dry

density

is a

m o r e

appropriate measure. Where density measurements are used in this

way a high dry density is usual ly sought .

Al though

high density is

not

of itself an important characteristic it implies that

oíher

properties

of the

soil

will

be

desirable

from

th e

engineering poiní

of

view.

A n increase in soil packing is accompanied by an increase in

sírength a decrease in

com pressibility

and a decrease in perm eability

which in

t u rn

can

lead

to

reduced

shrinkage/swell

problems.

Typical valúes

of

natu ral density

are

given

fo r

various

soil

types

in

Table 3.1.

T h r o ugho u í the chapter densi íy valúes ar e given in kg /m 3 ;

to

convert to unit weighís in kN/m

3

íhe mulíiplying factor is

0.009806.

For

g ranu la r

soils the

relative densi íy

is often

considered

to be

39



40

C O R R E L A T I O N S OF SOIL PROPERTIES

T a b J e 3 1 T Y P I C L V L Ú E S O F N T U R L D E N S I T Y

Natural

density

(kg/m

3

)

Material

Sands and gravéis: very

loóse

loóse

médium dense

dense

very

dense

Poorly-graded sands

Well-graded sands

Well-graded sand/gravel mi x t u r e s

Clays:

unconsol idated mu d s

soft,

open-síructured

typical,

normally

consolidaíed

boulder clays (overconsol idated)

Red tropical soils

Bulk density

1700 1800

1800 1900

1900 2100

2000 2200

2200 2300

1700 1900

1800 2300

1900 2300

1600 1700

1700 1900

1800 2200

2000 2400

1700 2100

Dry density

1300 1400

1400 1500

1500 1800

1700 2000

2000 2200

1300 1500

1400 2200

1500 2200

900 1100

1100 1400

1300 1900

1700 2200

1300 1800

1 Assumes

saturated

or nearly saturated condit ions.

more important than the absolute density. This is defíned as:

relative density

=

•

dr

— •

m x

m n

Par

Par,

where

p¿, p

dmax

and p

d m i n

a re the dry den sities in the fíeld and at the

densest and loosest síates of com paction

and

e,

e

max

and e

m

-

m

are the

corresponding void s ratios, respectively.

Because

of the difficulty of

measuring

fíeld

densities

in

sands

and

gravéis, valúes are usually estimaíed from standard peneíration test

results.

A

classifícation

of

relative densiíy

and SPT

iV-values,

although

widely used, has

received

repeated criticism.

Work by

Gibbs

and

Holtz

(1957) indicated that the relationship

beíween relative density and SPT valúes depends on the character-

istics of sand, w hether it is dry or saturated, and on íhe overburden

pressure. This

led to the

suggestion that correction factors

(C

N

) for

overburden pressure should be applied in the determination of

relative density and for f oun dation calculations.

Recommendations, from a

num b e r

o f

sources

are

given

in

Table

3.2. C orrected N valúes (A

r1

)

are

obtained using

th e formula:

=

CNJV

For

clarifícation

purposes ií should be noted that alího ugh the

interpretador

of

Terzaghi

and

Peck's (1948) classifícation, which

led



D E N S IT Y 41

ble

3 .2

S U M M R Y

O F

P U B L I S H E D C O R R E C T I O N F C T O R S

D f

Reference

~ f „ ,

orrection

factor

C N )

Units

of

overburden

l

¿OÍ4

K

Gibbs and Holtz 1957)

[equation by Teng 1962]

Peck and Bazaraa 1969)

Peck, Hanson and

Thornburn 1974)

ee 1976)

Tokimatsu and

Yoshimi

1983)

Q=

10 <

4

3 .25 0.5a;

2 0

C

N

= l-1.251og10cr;

1.7

^

0 7 a v

psi

ksf

kg/cm

2

or tsf

kg /cm 2 or tsf

n 2 or tsf

Liao and W h i íma n 1986)

kg/cm

2

or tsf

Skempton

1986)

C

N

=

1.7

0

For fine sands

of

médium

D r

For dense,

coarse sands

when normal ly

Consolidated

For overconsolidated

fine sands

kg/cm2

or tsf

to this particula r correction,

originated with

Gibbs and Ho ltz 1957),

the actual equation for the correction f acto r can be attributed to T eng

1962) .

A l t hough SPT correction fac tors we re discussed at some length by

Liao and W hitm an 1986), the

deímitive

work on the

subject

is that of

Skem pton 1986) . Ske m pton points

ou í

tha t

in

carrying

out the SP T

test the energy delivered to the sampler, and therefore the

blow

count

obíained in any given sand deposit at a particular effective over-

burd en pressure, can still vary to a

signifícan t

extent depending on íhe

m e t h o d

of releasing th e h a m m e r , on the type of

anvil

and on the



4 2 C O R R E L A T I O N S O F SOIL P R O P E R T I E S

ble

33 S U M M A R Y OF ROD E N E R G Y R A T I O S A F T E R S K E M P T O N 1986)

Hammer

Reléase

ER

ERJ60

Japan

China

USA

UK

D o n u t

D o n u t

Pilcon type

D o n u t

Safety

D o n u t

Pilcon Dando

oíd

standard

T o mb i

2 turns of

rope

Trip

M a n u a l

2 turns of rope

2

turns

of

rope

Trip

2

turns

of

rope

78

65

60

55

55

45

60

50

1.3

1.1

1.0

0.9

0.9

0 . 7 5

1.0

0.8

length of rods, if less than lOm. H is suggest ion is that N valúes

measured by any particular method should be normalised to some

standard

rod energy

raíio

ER

T

), and a

valué

of 60 is

proposed.

A

summary of rod energy ratios for a range of h a m m e rs and reléase

methods (wiíh

ro d lengths >

l O m )

is

given

in

Table 3.3 . N valúes

measured w iíh

a

k n o w n

or

est imated

ER

T valué

can be

normalised

by

the conversión:

60

where A represents other correction factors detailed in Table 3.4.

Skempton (1986) síates thaí th e Terzaghi-Peck limits of blow

coun t

for

various grades

of

relative density,

as

enumerated

by

Gibbs

and Holtz , appear to be good average valúes for normally con-

solidated natural sand deposits, provided that blow counts are

corrected for ov erbu rden pressure

N1

) and norm alised to a 60 rod

energy ratio C/Vj )^ ) ,

se e

Table 3.5.

Table 3.4 A P P R O X I M A T E C O R R E C T I O N S (A) TO M E A S U R E D N

V A L Ú E S A F T E R S K E M P T O N 1986)

R od

lengíh:

>10m

6-1 O m

4-6m

3^m

Standard sampler

U S

sampler wiíhouí

liners

Borehole diameíer: 65-115rnm

150mm

200mm

1.0

t.95

0.85

0.7

1.0

1.2

1.0

1 . 0 5

1 . 1 5



DENSITY 43

Table 3.5 T E R Z A G H I A N D P E C K S C L A S S I F IC A T T O N * ( A F T E R

S K E M P T O N

1986)

D

t

0 15

0 35

0.5

0 65

0 85

1.0

lassification

V e r y loóse

Loóse

é ium

Dense

Ver y dens e

NK-0.75

4

10

( 1 8 )

30

50

( 7 0 )

4 4

1 1

20

33

55

77

Ní 60

3

g

15

25

42

58

NiW

65

60

59

58

58

*CW=U; £Rr/

Another correction

often

applied to SP T valúe s wh en assessing the

relative density

of silts and fine

sands below

th e

water table

is :

with no correction for N v alúes of less tha n 15. This is based o n the

work of Terzaghi and it is suggested that, because of the lo w

permeability

of

such soils , pore water pressures

build up

d u r i n g

driving of the sam pler, resulting in increased . /V - valúes. This ap proach

is rec o mmen d ed by Tom linson 1980) in his discussion of the

application of corrections to SPT JV-values.

However, corrections

appear

to be somew hat academic in the

l ig ht

of

errors that can arise as a result of bad practice when carrying out

tests below the water table. In order to obtain

meaningful

resulís, the

borehole should be kept surcharged

with

wa ter a b o v e th e g ro u n d

wa ter

level

at all times. This is

often

neglected, both because it

requires

a

large supply

of

water

and

simply

out of

ignorance.

Consequently, groundwater

flows

into th e borehole, loosening th e

sand and resulting in

artificially

lo w

JV -values.

A lternatively, unrealis -

íically

h igh N - value s

may be obíained if drillers drive th e casing

ah ead of the borehole, to reduce th e problem of sand washing up the

casing, thus compacting

th e

sand beneath.

3 2 COM PACTED DENSITY

3.2.1 Compaction

test

standards

T he compacted density of a

soil

is not a

f undam e nt a l

p r o p e r t y but

de pe nds on íhe m a n n e r in which compaction is carried o u t .



44

C O R R E L A T I O N S OF SOIL P R O P E R T I E S

Compaction tests provide a s tandard m ethod of compact ion and a

standard

a m o u n t

of compacíive efíbrí to produce a soil density

against

which

site

valúes

can be com pared .

Soil is usually contained in a

mou ld

and compacíed using a

h a m m e r which

is

repeatedly raised

and

a l lowed

to

fall. Typical

compact ion equ ipment

is

illustrated

in

Figure 3.1.

To

cont ro l

íhe

compactive

effbrt -

the energy

per

uni t volume

- the

dimensions of

the

m ould and ram m er are precisely specifíed and the num be r of layers in

which

com paction is carried ou t , t he num ber of

b lows

per layer and

the height of

fall

of the ram m er are

al l

controlled. T here are basically

tw o

s tandards of com pactive eífort, commonly referred to as

stan-

dard

and heavy in the U .K. In the U.S. these are referred to as

s tandard and

modified

and are

detailed

in

A STM-D698/AA SHTO

T-99

and

ASTM-D 1557/AASHTO T-180, respecíively. Most tests

use

a special mould of abou t 1 litre capacity but for coarse-grained

soiís the larger California Bearing Ratio (CBR) mould is

used.

Slighí

s

oll r

ould

se

Rammer

1

V

es

Figure 3 1 Typical compaction

mould

and hand

rammer used incompaction tests



D E N S IT Y

45

Table

3 6

C O M P A R I S O N

O F E Q U I P M E N T SIZES, N U M B E R O F

R A M M E R

B L O W S A ND

N U M B E R

O F

L A Y E R S

O F

SOIL USED

IN V A R I O U S

C O M P A C T I O N T ES TS . D I M E N S I O N S

d , f A N D h A N D

WEIGHT W ARE SHOWN IN FIGURE 3.1

Test

designarían

BS 1377:1975

Test

12

Test 12 (modified)

Test

13

Test 13 (m odified )

AASHTO

T145

TI

80

TI 80 (m odified)

Mould

volume

d

1.0

2.32

1.0

2.32

0.94

0.94

2.32

Mould

día d

( m m )

105

152

105

152

101.5

101.5

152

Mould

ht h

( mm)

1 1 5 . 5

127

1 1 5 . 5

127

1 1 6 . 4

116.4

127

Rammer

wt W

k g )

2.5

2.5

4.5

4.5

2.50

4.54

4.54

R a m m e r

ll

m m )

300

300

45 0

450

304.8

457.2

457.2

Number

of

layers

3

3

5

5

3

5

5

Blows

per

layer

27

62

27

62

25

25

56

The modified forras of the test use a CBR mould and are su i t ab le fo r coarser

soils.

differences exist between British and Ame r i can

S tandards ,

as in-

dicated

in

Table 3.6, which g ives mould

and

rammer s izes

for the

var ious

tests.

With sands a nd gravéis, th e ramm er tends to displace th e mater ia l

ra the r

t han

compací i t so that the densities obtained in the

compaction test

a re

unr ealisíically low when co mpa red wi th wha t

ca n

be achieved on site. To overeóme this, a v ibra t ing h am m e r can be

used

instead

of the

r am m er . V ib r a t io n

is

typically carried out

for 60

seconds

per

layer

und e r a

constant

forcé o f

30 40kg.

3.2.2 Typical

compacíed densities

The com pacted den sity achieved for a soil depends on the soil

type ,

its

mois íure

contení

and the compactive effort used. Table 3.7 shows

typical valúes of máximum

d ry

densi ty (MDD) and optimum

moisture coníení

fo r

soil classes, using

íh e

Unified classifícation

sysíem,for soils

compacíed

to A A SH TO or BS s tandard

compaction:

A A S H T O

T99

(5.51b

r a m m e r m e t h o d )

or BS

1377:1975 Test

12

(2.5kg ram m er m ethod) . The valúes given are b ased on typical valúes

given by Krebs and Walker (1971) and the U.S. A rm y Engineer

W a íe rways

Exper iment

S ta t ion (1960),

and on the

a u tho r s

ow n

experience. A similar set of valúes but r elated ío íhe A A SHT O soil

classifícaíion system, is given in Table 3.8. These a re based on íhe

above valúes

and the

re la t ionship between

the

A A SH TO

and

U nified

soil classifícation systems, and on va lúes sugge sted by G regg (1960).

I t should be noted that clean sands often show no clear o p t imum



4 6 CO RRELA TI O N S OF SOIL PRO PERTI ES

Table

3 7

T Y P I C A L C O M P A C T E D D E N S I T I E S

A N D

O P T I M U M M O I S T U R E C O N T E N T S

F O R

S O I L

TYPES

USING

THE UNIFIED CLASSIFICATION SYSTEM

Soil

description

Gravel/sand mixtures:

well-graded, clean

poorly-graded, clean

well-graded, small silí

content

well-graded, sm all clay co ntent

Sands

and

sandy

soils:

well-graded, clean

poorly-graded, small silt content

well-graded, small silt contení

well-graded, small clay content

Fine-grained soils o f l o w p last icity:

silís

clays

organic silís

Fine-grained soils of high plasticity:

silts

clays

organic clays

Class

GW

GP

G M

GC

SW

SP

SM

se

M L

CL

OL

MH

CH

OH

M D D

standard

compaction

(kg/m

3

)

2000 2150

1850-2000

1900-2150

1850-2000

1750-2100

1600-1900

1750-2000

1700-2000

1500-1900

1500-1900

1300-1600

1100-1500

1300-1700

1050-1600

Optimum

moisture

content

( )

11-8

14-11

12-8

14-9

16-9

21-12

16-11

19-11

24-12

24-12

33-21

40-24

36-19

45-21

Table 3 8 T Y P I C A L C O M P A C T E D D E N S IT IE S AND O P T IM U M M O I S T U R E C O N T E N T S FOR S O I L

TYPES USING THE AASHTO SOIL CLASSIFICATION SYSTEM

Soil description

Well-graded gravel/sand mixtures

Silty or

clayey gravel

and

sand

Poorly-graded sands

Silíy

sands

an d

gravéis

of

lo w

plasíicity

Elastic

silts diatomaceous or micaceous

Plástic

clay, sandy clay

Highly plasíic or elastic clay

Class

A -l

A-2

A-3

A-4

A-5

A-6

A-7

BSIAASHTO

Max dry

densiíy

(kg/m

3

)

1850-2150 •

1750-2150

1600-1900

1500-2000

1350-1600

1500-1900

1300-1850

compaction

p í moisture

contení

( )

5-15

9-18

5-12

10-20

20-35

10-30

15-35

moisture content and that peak densiíy may be achieved when íhe

sand

is

completely dry.

Work carried out by Morin and Todor (1977) on red tropical soils

;orrelations betvveen the opt imurn

n

África

and Sou th Am erica gave



DE NS I T Y 47

4

Plástic limit -

(a)

Opíimum

moisture

contení -

(b)

Figure 3.2 Relationships of optimum moisíure contení wiíh

plástic

limií and with

áx u

dr y

density

for red

tropical soils

after

Morin

and Todor,

1977)

mois ture conten í

and

plasíic limií

and

be íween opí imum mois íure

conten í a nd m á x i m u m d ry densi íy as indicated in Figu re 3.2. M orin

and odor also produced a relaíionship beíween opíimum mo isíure

coníent

and íhe

perceníage

of

paríicles

f íner than 2¿um buí

íhis

showed too wide a scal ter to be of use and has noí been included.



8 CORR ELATIONS OF SOIL PROPER TIES

1 55

6 8 10 12 U 16 18 20 22 24 26 28 30 32 34 36 38 40

Moisture

contení - of dry

w ight

Figure 3.3 Typical moisture-densüy curves modified

after

Woods and Liíehiser, 1938

and Joslin, 1959)



DENSITY 49

3 3

Typical moisture density curves

Work

carried out by

W oods

and

Litehiser 1938)

in

Ohio indicated

tha t , fo r Ohio soils, nearly al l m oisture-d ensi ty curves

have

a

characteristic shape. On the basis of o ver 10,000 tests 26 typical

curves

were produced,

as

shown

in

Figure 3.3.

Use of the

curves

allows the m áxim um dry densi ty and op t imum m oisture content to be

estimated f rom a single po int on the curve, greatly reducing time and

eífort.

I t should be noted that the curves are plots of bulk densi ty ,

instead of the more usual dry density, against moisture content. The

inset

table gives íhe corresponding m áx im um dry density and

optimum moisture content

for

each curve. When used with rapid

mois ture

content determinations,

these

curves provide quick and

fairly accurate estimates. They have been found to be applicable in

many

áreas though minor

modifications

have sometimes been

necessary. Accuracy

is

improved

if the

moisture content

of the

test

specimen

is

cióse

to

optimum

and

preferably

o n the dry

rather

than

the wet side. The curves are not valid for unusual materials such as

uniformly graded sand, highly micaceous soils, diatomaceou s earth,

volcanic

soils

or

soils

in

which

the specific

gravity

of the

solids

diífers

greatly from 2.67.



Chapter

PERMEABILITY

The coefncient of permeabil i ty is defíned as the quan t i t y of flow

through uni t área o f soil un de r a unit pressure g radient. T his assum es

a l inear reíationship between the pressure gradient and quan t i t y o f

f low, q,

which

is the

basis

fo r

Darcy s

l a w :

4J

where k is the coefficient of perm eabií i ty

A

is the

área

of flow

and i is the hyd raulic pressure gradient.

If the vo lume of f low q is divided by the área A

then

the veloc ity of flow

v i s obíained and Equa t ion (4.1) can be w ri t ten :

*-?

i

(4.2)

From this,

it can be

seen thaí

th e c oefficient of

perm eabil i ty

can be

thought

of as the

veíociíy

of flow

that results f rom

a

unit pressure

gradient. Since pressure is usually measured as head o f w ater an d

pressure

is

loss

o f

head

per

unit distance, i typically

has the

dimensions m/m so thaí

k

has the units of veíociíy; typically m/s

Ho wev er, i í sho uld be remem bered that área A is the to tal

área

of soil

being considered

but

parí

o f

íhis

área

will

be

occupied

by solid

partióles so íhe

área

of flow wilí be less. This means íhaí veíociíy u is

only a no íional valué, used for calculaí ing vo lum es of f low, and íhe

true average veíociíy of flow u will be greater:

l

e

n

where

e

an d

n

are the

vo ids rat io

and

porosi ty

of íhe

so il, respectively.

T he

permeabil i ty

o f a

soil

is

s írongly iníluenced

by its

mac r o-

50



P E R M E A B I L I T Y 51

s t r uc t u r e : clays coníaining físsures or fine bands of sand will have

permeabilities which are many times that of the clay material itself.

Also,

since

flow

tends

to follow th e

line

of

least resisíance, stratiñ ed

soils

often

have horizontal permeabilit ies which are m a n y times the

vertical permeability and the overall perm eabil ity

will

be approxi-

mately equal to the horizontal permeability.

Because

of the sm all size

of labor atory specimens

and the w ay they are

obtained

and

prepared,

large-scale

features are absent and test results do not

give

a t rue

indication

of fíeld

valúes

in

soils w ith

a

pronounced

macro-structure.

Moreover,

laboratory tests usually constrain w ater

to flow

vertically

th rough

the specimen whereas the horizontal permeability

m ay

be

much greater , and henee of

overriding

importance so far as

site

conditions

are

concerned. Field tests overeóme

these shortcoming,

bu t , since íhe pattern of w a ter flow from a well can

only

be guessed,

iníerpretation of íhe test results is

diííícuíí

and uncerta in. Thus, one

set of

problems

is exchanged fo r

another

4 1 TYPICAL VALÚES

The íypical range of valúes encounfered is indicaíed by

Table

4.1,

which is based on

informalion

originally presented by Casagrande

and

Fad um 1940). Superimposed

on íhe

charí

are

íypical valúes

fo r

compacíed soils,

classifíed by íhe U nifíed

sysíem. These

re late to

soils

compacíed using

the heavy

compaction

slandard:

AASHTO

T-180

lOlb r a mmer ) or BS 1377:1975, Tesí 13 4.5kg rammer). Typical

permeabiliíy valúes for highw ay matería ls , suggested by Krebs and

Walker 1971),

are

given

in

Table 4.2. A ddiíional

informaíion on the

influence of voids ratio in differení soil types is given by Mitcheíl

1976).

4 2 PERMEABILITY AND GRADING

A

theoreíical

equaíion relaíing the coefíícient of permeability ío íh e

soil and permeaní properíies was

developed

by

Táylor 1948). This

gave:

e

wfaere

k

is the

coefficient

of perme ability

D

s is

some effective paríicle diameíer



Table

4 .1

T Y P I C A L

P E R M E A B I L IT Y V A L Ú E S

F O R

S O I L S

Coefficient of

permeab i l i ty

lo g scale)

Drainage

c o n d i t i o n s :

Typical soil

g roups:

10

II 0-10 1Q 9

1 8 io - 7

i .

10

6

I

i o 5

I

m /s

109

10

10

7

1 0

10

1 0

cm/s

10

1

f t / s

io-

3

i

10

2

9

10 10

6

10 ~ 5

10

10

MH

MC-CL

Practically

i mp e r me a b l e

Very low

L ow

M é d i u m

Pract ica l ly

im per m eab l e

Poor

G C — • G M —

CH SC

S M

SM-SC

SW-K

SP->

Soil types:

H o m o g en eo u s

c lays

be low

the zone of

w e a t h e r i n g

Silts,

fine sands, sil ty sands ,

glacial t i l l , strat i fied c lays

Cl ean

sand

a n d

grave l

Fissured and weathered clays and c lays

modified by the eflects of vege ta t ion

Note:

th e

rrow dj cent

lo



P E R M E A B I L I T Y

53

ble

4 2

T Y P I C A L P E R M E A B I L I T Y V A L Ú E S

F O R H I G H W A Y

M A T E R I A L S

Material

Permeability

(m/s)

niformly graded coarse aggregate 0 .4-4

x 10 ~

3

Well-graded aggregate w i t h o u t fines 4x 10~3-4x 1 0 ~

5

Concrete sand,

lo w

dust content

7x 10~

4

-7x

1 0 ~

6

Concrete

sand, high dus t c on te nt 7 x l O ~6 - 7 x l O ~ 8

Silty

an d

clayey

sands

10~7-10~9

Compactad silí 7x 10 8-7x 1 0 ~1 0

Compacted clay less tha n 1 0 ~9

Bi t u mi nou s concrete*

4x

10~5-4x 10 ~8

Portland cemen t concr ete less th an 1 0 ~1 0

* New pavements ; va lúes as lo w as

1 0 ~1 0

have been reported fo r sealed,

traf l íc-compacted

h ig hw a y p a v e m e n t .

y is the uni t we igh t or weight density of the permeant

\i

is the

viscosity

of the

permeant

e is the void s ratio

and c is a shape factor.

In soils, the permeant is usual ly water and the efíective particle

d iamete r D s

is

usually taken

as the 10 (or eífective)

particle size

D

10

.

Yhis

led to the

Hazen fo rmula :

y

e3

where the constant C, repíaces - —

Based

on

experimental work wi th

clean

sands, Hazen (1911)

proposed a valué of betw een 0.01 and 0.015 for C15 where k is in m/s

and Z > 1 0 is in m m . However, this ignores the

large

efíect that even

small

changes

in

e will have

on the

valué

of

k

as can be

seen from

Taylor's

equat ion, and can be expected to give only very appro ximate

resuíts. For instance, experimental work by Lañe and W a s h b u r n

(1946), reporíed in Lambe and Wh itma n (1979) gives

l

valúes of

beíween 0.01

and

0.42 wit h

an

average valué

of

0.16, w hils í Hol tz

and

Kovacs (1981) suggesí a range of 0.004 ío 0.12 with an average valué

of 0.01.

The

equat ion

is

usu aíly considered

ío be

valid

for

soils hav ing

a coefficient of permeability of at

least

10~

5

m/s.

Figure 4.1 gives ploís of k againsí D

10

, based on experimental

results , in w hich the valué

o í e

has been taken into acc ou nt. It

will

be

noted

tha t

the

correlaíions given all relate

to

sands

and

gravéis.

T he

greaíer ran ge of particle size wh ich is present in m ost clays and íhe

effecís

of the clay mineralogy m ake such correlations m ore resíricíive

fo r

clays. Some useful

informat ion

on the permeabil i íy of clays is

provided

by Tavenas et al. (1983a and b),



5 CORRELATIONS

OF SOIL PROPERTIES

0.05

w

X

E

o.01

O.OO5

o

c

o

© O OO1

o

ü

O O O O 5

O O O O 1

Burmister

Cu= 1.5, e = 0.75

Cu

= 3, e = 0.7

Mansur

Mississippi

ríver

sands

Cu

=2 - 3

e = 0.9 - 0.6 ,

- field tests

- Icb

tests,

Í

V

Hazen

formula

Limited to D-0= 0.1 — 3mm,

Cu<5

USNavy

Correlation oí lab test valúes

of various materials

Cu

= 2 — 1 2

íower

Cu valúes are

associated with higher e vaiues )

Liirited to D10/Dg less than

1.4

D1O/DS>1.4 cr C

u

?12

lie in a

tange

of

tower permeabilities

NOTE: correlations

shown are for remolded

compacted

sands and sand-grave mixtures

Cu = cc«í í ¡cieni oí

ufiiíOírnity

e

= vo ids r a t i o

O.1

0 5 1

Grain size, D10 -

mm

10

Figure 4 1 The

permeability

of sands and

gravéis



Chapter

5

CONSOLID TION ND

SETTLEMENT

The

settlement

of

soils

in

response

to

loading

can be

broadly

divided

into

tw o

types: elastic settlement

and

time-dependent settlement.

Elastic

settlements

are the simplest to

deal

with; they

are

instan-

taneous, recoverable,

and can be

calculated

from

linear elastic theory .

Time-dependent settlements

occur in both

granular

and

cohesive

soils,

although

the response time for gran ular soils is usua lly sh ort. In

addition

to

being time-dependent, their response

to

loading

is

non-linear

and deformations are only partially

recoverable.

Two

types of time-dependent settlement are recognised. Primary consoli-

dation results from the

squeezing

out of water from the soil voids

under

th e influence of

excess pore w ate r pressures, g ene rated

by the

applied loading. Secondary compression occurs essentially

after

all

the

excess

pore

pressures have been

dissipated

that

is ,

after

primary

consolidation

is

substantially com plete,

but the

mechanisms

involved

are not fully understood. The

settlement

of

granular soils

is

more

difficult to

predict with

any

accu racy, largely because

o f the difficulty

of

obtaining and testing und isturb ed soil samples, and settlements are

usually estimated

by

indirect methods.

Alteraatively,

píate

bearing

tests

m ay be

used

but

their results

are difíicult to

interpret.

5 1

COMPRESSIBILITY

OF CLAYS

The compressibility of

clays

is

usually

measured by means of

oedometer consolidom eter) tests,

or

similar method s see Tave nas

and

Leroueil 1987). Results

may be

expressed

in a

number

of ways,

leading

to a,

sometimes

conftising,

variety

of

compressibility

par-

ameters. As indicated in F igure 5.1, either am pie thickn ess, h or voids

ratio

e may be

plotted

againsí

consolidation

pressure, p which may

itself be

plotted either

ío a natural scale or,

more usually,

to a

logarithmic

scale.

55



56

CORRELATIONS OF SOÍL

PROPERTIES

V i r g i n

c omp r e s s i o n c u r v e

O

O

2 4 6 8 1O

o

Consolldation prossur* , p MN/m

a)

OverconsoJidation

pressure

= C

b.

O

Unloading

Recompression

O 01 O t

1 10

Consoüdation prsssura, p - MN/m

í

Figure

5 1

Typical

ploís of compressibiliíy t st

results

5.1.1 The compressibility parameíers

The process of compression on a soil can be usefully ill-.otrated by

means of íhe

model soil sample

as illusírated in Pigure 5.2.

Recognising thaí compression íakes

place by a

reduction

in

the

volume of voids with virtually no change in íhe volume of íhe solid

paríicles compressibiliíy was originally defíned by íhe eoeffkie í of

compressibiliíy

a

which

is íhe

change

in

voids

ratio per

unií increase



CONSOLIDATION A ND SETTLEMENT 57

Pressure p1

l l lí

Voids

Solías

Pressure

Pffdp

=

de

Vol e

Vol 1

Yoids

oii s

dh

Figure 5.2 Compression of the

model soil

sample

in

pressure.

In terms of the

model soil

sample,

de e

e >

y ~™~

P2~Pi

5.1)

and is the

slope

of the

curve shown

in Figure

5.1

a)

when e

is

plotted

against p . From an

engineering

viewpoint, it is the proportional

change of thickness of a specimen

that

is of direct concern. For a

constant

cross-sectional área

this is

proportional

to the

proportional

change

of volume of a soil an d gives rise to the concept of the

coefíiclenl

of

volunie

of

compressibility

m

v

,

which is much more

commonly used:

d volume) 1 dh 1

v volunie dp h áp

Refemng

to the soil sample, m

v

can also be expressed in terms of the

voids

ratio:

dh 1

1

5.3)

This is the slope of íhe curv e in Figure 5.1 a)

when

h

is plotted against

p.

From

Equations

5.1 and

5.3,

th e

relation ship between these

tw o

deímitions of

com pressibility

is :

av

=

my l+e) 5.4)

It can be

seen thaí

the

slope

of the

curve

in

Figure

5.1 a) is not

con stant. This m eans that

th e coefficients

a

v an d

m

v

also var y

and

that

a

given

valué applies only to a

specific

pressure range. However, the

curve obtained in

figure

5.1 b) when the logarithm of consolidation

pressure is used, approxim aíes m uch m ore closely ío a

straight line,

at



5 8 C O R R E L A T I O N S

O F

SOIL

P R O P E R T I E S

least

on the

v irgin comp ression curv e. This gives

rise to two fur ther

measures of compressibility, the

compression

índex, C

c

, and th e

modifíed

compression

índex or

compression ratio,

CC£, which are the

slopes of the

virgin compression cu rves obtained

by

plotting

e

or

h,

respectively, against

l o g p :

áe

dh

d logp) logpa-logp logípa/pi

de 1 e -e 1

C

C£

-

-

T

/d logp)-

~

1 Iog p

2

/

Pl

)

5.5)

5.6)

Note

that,

for these

evaluations, logarithms

are

taken

to the

base

10. From equations 5.5 and 5.6, íhe relationship between C

c

and C

ce

foliows that between

a

v

and

m

v

:

C^CJl+eJ 5.7)

O f

th e

two,

c is

much m ore comm only used. From

equations 5.3 and

5.5, it can be relaíed to mv:

1

v

givmg

5.8)

For the

com pression parí

of the

curve,

the

terms

recompression

índex, ,

C

r5

and modiíled recompresslon Index,

C

r£

, are used,

defined

in the

same ways as

Cc

and C

C£

, respectively.

5.1.2

Setíleinení calcóla

tions

using consolida

tion theory

Returning

to íhe

basic defíniíion

of the

coefficient

of volume

compressibility, given

in

equ ation 5.2:

áh 1

iri

h áp

5.9)

li can be

seen t ha t, once

my is

k n o w n

for a

particular pressure range,

th e compression,

dh,

of a

layer

of íhickness, h, due to a

load

increment, dp,

can be calculated by simply

íurning

th e above

equation

a round:

áh



CONSOLIDATION A N D SETTLEMENT 59

since dh is normally thoughí of as íhe setílement, p and áh is the

applied pressure increase,

< j

this becomes:

p

= Ham,

(5.10)

where

specimen íhickness,

h

is now replaced by íhickness,

H

of

íhe

compressible síraíum. The

average valué

of a

across

a

compressible

layer, due lo some applied loading, is usually calculaíed using

elaslicity theory. Allhough nol strictly valid

fo r soils, ií gives

sufficienlly

acc uraíe valúes. Selílemenl

is Ihen

oblained using consoli-

daíion theo ry by w ay of Eq uatio n 5.10.

W here valúes o f

C

c are obtained, m

v

valúes may be calculated from

Equation 5.8, using

th e appropriale

valúes

of

consolidaíion pressure

and

voids

ratio. Alternaíively, Equalions 5.8 and

5.10

may be

combined and seíílemenl calculated direcíly from

C

c valúes:

Iog(p 2/Pi)

l e p-

givmg

= C

l e

5.1.3 Settleoiení

calculations

using elasticity theory

An alterna

ti ve approa ch is to calcúlate d isplacem ents (seíílements)

directly using elasticiíy the ory , thus reducing

t h e tw o

sepárate stages

in

th e

seítlement calculation

ío one, and

obviaíing

the

need

to

calcúlate average valúes

of

consolidaíion pressure across

soil layers.

Numerous solutions, for

both

síresses and displacements, have been

produced,

man y

of wh ich ha ve been

presented

by

Poulos

and Da

vis

(1974).

The

problem

wiíh

usin g elastic soluíions

ío

calculaíe seíílernenís

is

thaí ií requires the evaluation of Young's m odulus , E and Poisson s

raíio, v, neither of which are

measured,

or are strictly meaningful, fo r

soil

consolidaíion

problems. Considering

Equation

5.9, since

the

raíio áh/h can be

íhoughí

of as a

sírain,

m y is sírain/síress,

w iíh units

I/stress; íypically

m

2

/kN

or m2/MN. Thus, ií is by defíniíion akin ío

íh e reciprocal of Young's modulus , £, and whereas E can be

envisaged simplisíically as íhe síress req uired ío do uble íhe length o f

an object, mv can be envisaged as an área of soil wh ich,

if

subjecíed to

a

unit load,

will

jus t disappear

Of course,

such absurdiíies

do noí

occur

in

realiíy because

íhe relationships are not

valid

fo r íhese

extremes. Addiíionally,

íhe

relationship beíween E

and

mv

is not a

simple

reciprocal

one because E is defíned for a specirnen wiíh



60

CORRELATIONS OF SOIL PROPERTIES

unrestrained sides whereas m

v

is

definH

for a

specimen which

is

laterally constrained. The relationship

bt

ween and

m

y therefore

depends on the valué of Poisson s ratio, t h \ . > £ \

1 (l + v ) ( l- 2 v )

-

This

relationship can then be used when calcúlate lg settlem ents

using

elastic

theory.

When used in this context, is nc>

^strictly

an

elastic constant,

but it

does represent

the

response

of

thc soil

to a

single loading applied over a long

period.

To emphasise the p¿*nt, the

term deform ation m odulus is sometimes used for defined L this

way. Thus, eíastic theory

can be

used

to

calcúlate

consolidaron

settlem ents, even thoug h these are not elastic (i.e. recoverable). T¿ 7

main problem lies in obtaining a valué of Poisson s ratio that

properly represents the consolidation behaviour of soils. Poisson s

ratio is not m easured in standard soil testing an d, indeed, it is

virtually im possible to obtain realistic m easure m ents. Howev er, it

has been po inted out by Skem pton and Bjerrum (1 957) that very little

lateral

strain

occurs

during the

consolidaíion

of clays so

that,

efíectively, Poisson s ratio is zero, and

•-,-;—a

where M is the defonnaíion m odulus or constrained m odulus.

Another reason

fo r

choosing

a

zero valué

is

that calculated

seítlem ents based on elastic solutions then becom e identical w iíh

those

based on consolidaíion íheory, which has been shown over the

years to give reasonable predictions provided that suitable correc-

tions are

m ade

for the

pore pressure response

o f the

soil (Skem pton

and Bjerrum 1957).

5 1 4 Typical

v alaes and

correlatioos

of eom pressibiüty

eoeffkients

Typical valúes of the coefficient of

volumc

com pressibiliíy, mv are

indicated in

Table

5.1, along with

descripti

?

íerms

for the

various

ranges of com pressibility. Althoug h m y is the m ost suiíable, and m ost

popular, of the

com pressibility coefficients

for the

direct

calculation

of settlem ents,

its

variabiiity with

confining

pressure m akes

it

less

useful

when

quoting typical conipressibilities

or

when correlating

compressibñity with some other property. For íhis reason, the



CONSOLIDATION AND SETTLEMENT 61

Table

5 1

T Y P I C L

V L Ú E S

O F T H E

C O E F F I C I E N T

O F

V O L U M E C O M P R E S S I B I L I T Y

A N D

DES CRIP TIVE

T E R M S

U S E D

A F T E R C Á R T E R 1 98 3)

Type o f clay

Descriptive

t erm

Coefficient o fvolume

compressíbili ty, /nv

m

2

/MN)

f t

2

/ t o n )

H e a v y

over-consolidated

boulder

clays, stiff weathered rocks e.g.

weathered mudstone)

and hard

clays

Boulder clays, marls, very stiff tropical

red

clays

Firm

clays, glacial outwash clays,

lake

deposits, weathered marls, firm boulder

clays,

normally

Consolidated clays

at

depth and firm tropical red clays

N ormally Consolidated alluv ial clays

such as estuarine and delta deposits,

and sensitivo

clays

Highly organic alluvial clays and peats

Very lo w

compressibility

L ow

compressibility

M é d i u m

compressibility

High

compressibili ty

Very

high

compressibility

<0.05

0.3-1.5

< 0.005

0.05-0.1

0.005-0.01

0.1-0.3 0.01-0.03

0.03-0.15

>0.15

Table

5.2

1981)

T Y P IC A L V A L Ú E S

O F

C O M P R E SS IB IL IT Y I N D E X , Cc A F T E R H O L T Z

A N D

K O V A C S

Soil

Normally Consolidated médium sensitivo clays

Chicago silty clay CL)

Boston blue clay CL)

Vicksburg Bucksho t clay CH )

Swedish médium sensitive clays

CL-CH)

Canadian

Leda clays

CL-CH)

México City clay M H )

O rganic clays O H )

Peats Pí)

Organic silt

and

clayey

silts

ML-MH)

San Francisco Bay Mud

CL)

Sa n Francisco Oíd Bay clays CH )

Bangkok clay CH )

0.2 to 0.5

0.15 to 0.3

0.3 to 0.5

0.5 to 0.6

1

t o3

1 to4

7to 1 0

4 and up

10tol5

1.5 to 4.0

0.4 to 1.2

0.7 ío 0.9

0.4

compression Ín dex,

C

c

,

is

usually preferred. T ypical v alué

of

com pres-

sion

índex

are

given

in Table 5.2.

Skempton 1944) proposed th e

folio wi n g

relationship between

compression

índex

and liquid limit

LL)

fo r normally-consolidated



62 CORRELATIONS OF

SOIL

PROPERTIES

Table 5 3 S O M E P U B L I S H E D C O R R E L A T I O N S F O R C O M P R E S S IO N Í N D I C E S A F T E R A Z O U Z E T

A L .

1976)

Equation

Regions of

applicability

Cc=0.007

(LL-7)

Ce,=0.208e0+0.0083

Cc

= 17. 66xKT5 > v j

Cc=1.15(e0-0.35)

Cc=0.30(e0-0.27)

=

l.15x10

2

Cc

= 0.75(e0-0.50)

€« = 0.1566

C = O . O l H >

3 wn-1.35x10

-1

Remoulded

clays

Chicago

clays

Chicago clays

All clays

Inorganic,

cohesive

soil; silt,

some clay; silty clay; clay

Organic

soils-meadow

mats,

peats, and organic silt and clay

Soils of very low

plasticity

All clays

Chicago

clays

As

summarised by A zzouz, Krizek, and Corotis (1976).

Note: w0

=

natural water

contení.

clays:

C =

0.007(LL-10).

Terzaghi and Peck (1967) proposed a similar relationship, based on

research with clays of low and médium sensitivity:

CC = 0.009(LL-10).

This

relationship

has a

reliability range

of

+30

and is valid for

inorganic clays

of

sensitivity

up to 4

(see Chap ter

6) and

liquid

limit

up to 100. Based on the work of Skempton and Northey (1952) and

Roscoe et al (1958), W ro th and Wood (1978) used critical sta te soil

niechanics

considerations

to deduce a relationship between cornpres-

sion

índex and

plasticity índex (PI)

for

remoulded clays:

where

Gs is the

specific

gravity of the soil solids. Table 5.3 produced by

Azzouz

et al (1976) gives

a

summary

of a

number

of

published

correlations.

The recompression índex, C

r

, is

defined

in the same way as C

c

except tha t it applies to the unlo,?ding phase of the cons M idation test.

Typical valúes of

C

r

range from 0)15 to 0 .35 (Roscoe ei

¡ I

1958) and

are often

assumed

to be

5-10

of

Cc.

5 5 ettlement

corrections

If the results of oedometer tests are used directly to calcúlate

settlements, the valúes obtained tend to ov er-estimate the settlements



CONSOLIDATION

AND

SETTLEMENT

63

that actually occur, particularly wi th

overconsolidated

clays. An

exception

to

this

is in the

case

of very

se nsitive clays, wh ere predicted

settlements m ay slightly und er-estim ate actual valúe s. Th e reason for

this is that the p ore pressure response of ciays in the fíeld differs from

that of

confined laboratory specimens. This

has

been discussed

by

Skempton

and

Bjerrum (1957),

w ho

show that

th e

ratio

of

actual

settlement to calculated settlement depends on both th e response of

th e

pore water pressures

to

applied

loads and the

geom etry

of

each

problem.

The

response

of the

pore w ater pressures

to

loading

can be

measured in the triaxial test and is expressed in terms of Skem pton s

(1954) pore pressure

parameters A

and

B

For saturated clays, actual

settlement

p f ie id, is given by:

h«av¡ly ovar- over -

een so I ¡ di

f d onsolidated normally

•nd? clays

clayí consolida t d

clays

clay

1 2

0 2 0 4 0 6 0 8 1 0 1 2

Pore pressure

coefficient,

Figure

5.3

Typical

valúes of the factor

\ifor afoundaíion

width b on a compressible

layer of

thickness

h afíer

Skempton, 1954)



Table 5 4 T Y P I C A L V A L Ú E S

O F

C O N S O L I D A T I O N F A C T O R

n F O R

V A R I O U S T Y P E S

O F

S O IL A T E R C Á R T E R 1983

Ty pe o f clay

Definiti

=

0 5

H b=l

V e r y

sensitive clays

soft al luvial ,

e s t u a r i n e , ma r i n e c l a y s )

N o r m a l l y

Consol idated c lays

O v e r -con so l id ate d clav

Lias,

L o n d o n , O x f o r H , ,i ld clays)

H e a v i l y over-consol idaí v-J

clays

B o uld e r c lay , m ar l )

1.0-1.1 1.0-1.1 1.0-1.1

0.8-1.0 0.7-1.0 0.7-1.0

0.6-0.8

0.5-0.7 0.4-0.7

0.5-0.6 0.4-0.5 0.2-0.4

1

1

j

b

tompresslble layer

Surface

la yer

•

•




where p is the

calculated oedometer settlement

and

¡

is a

factor which

depends on the pore pressure parameter.

The distr ibution of

stresses across

a

layer

of

soil depends

on the

ratio

of

w id th , b of

a foundat ion to

thickness, H

of the

layer. Valúes

of

^

can be

obtained

for

given valúes

o f

pore p ressure parameter,

A

from

Figure 5.3. Valúes

of

parameter

A are not

normally measured

in

the laboratory tests commonly used for foundation design but they

are found to

depend

on the

consolidation history

of the clay,

particularly

the

degree

of overconsolidation. For

most practical

purposes

it is suffícient to use

valúes

of \i

selected from Table 5.4.

5.2

RATE

OF CONSOLIDATION OF

CLAYS

The

rate

of

set t lement

of a

saturate d soil

is

expressed

by the

coefflcient

of

consolidation,

c

v

.

Theoretically, consolidation takes

an

infínitely

long time to be completed and it is usual to calcúlate the time taken

for a

given degree

of

cons olidation, U

to

occur, w here U

is defined

by :

Consolidation

settlement after a given time,

r

Final consolidation settlement

The

time,

í for a

g iven degree

of

consolidation

to

occur

is

given

b y:

where d is the máximum length of the drainage path

equal

to half

the

layer thickness

for

drainage

top and

bottom)

and 7 ^ , is called the basic time facto r. Valúes of T

v

for various valúes

of U are given in Table 5.5.

The

rate

of

settlement

of a

soil,

an d

henee

th e

valué

of cv, is

governed

by two

factors:

th e

amount

of

water

to be

squeezed

out of

th e

soil

and the rate a t

w hich that water

can flow

out.

The

amount

of

water to be squeezed out depends on the coeñlcient of compress-

ibility,

mv,

and the

rate

at

which

it

will

flow

depends

on the

coefficient

of

permeability,

k.

The relationship between

c

v

,

m

v

and

k

is :

m

v

* v /

W

w

where y

w

is the weight density unit weight) of water .

Because

of the

wide range

of

permeabilities that exist

in

soils,

th e

coefficient of

consolidation

can itself

vary widely, from

less

than

Im2/yr

for

clays

of low

permeability

to

1000m

2

/yr

or

more

for

very

sandy clays, fissured clays and w eathered rocks. Some typical valúes



A n y p ressure d i s t r i bu t ion ,

dra inage

t o p a nd

bo t tom

Decrea sing pressure, drain age

a t

bo t tom on ly

Table

r r

0.1

0.2

0.3

0.4

0.5

0.6

0.7

0.8

0.9

5.5 V A L Ú E S OF

TIME

Casel

0.008

0.031

0.071

0.126

0.197

0.287

0.403

0.567

0.848

T ,

Case

2

0.047

0.100

0.158

0.221

0.294

0.383

0.500

0.665

0.940

:

'

F A C T O R , Tv

• •

J

Drainage condit ions an d pressure distrib

Case

3

Casel Case

2

0003 . . - . - . . - . ; • - . . .

• •

: : • • •

: • : • • .

:: :•.••::••

- 6 ií 4 < « < > s í í? s X s a i « S ^ i » » < > i x .

0.009

0.024

0.048

0.092

0.160

0.271

0.440

0.720 ; . . ' • . i . ; . : . - . . ..:.'.... . . . - .

i - 1

..

,

. .

.

i- ¡•.•'.•'.-••V;'."..-.

• • ? . ' . ' •

* Case 1 m a y b e used fo r un i f orm pre s sure d i s t r ibut ion wi th d r a i n a g e a t top o r b o t t o m o n l y .



CONSOLIDATION

AND

SETTLEMENT

67

Table 5 6 T Y P I C A L V A L Ú E S O F T H E C O E F F IC I E N T O F C O N S O L I D A T IO N

v

Soil

cm2/sxl T4)

m2/yr)

Boston

blue

clay

CL)

Ladd and

Luscher, 1965)

Organic silt OH)

Lowe, Zaccheo, and Feldman, 1964)

Glacial lake clays CL)

Wallace

and

Otto, 1964)

Chicago

silty

clay CL)

Terzaghi and Peck, 1967)

Swedish médium

sensitive

clays CL-CH)

Holtz

and Broms, 1972)

1. laboratory

2. field

San Francisco B ay Mud CL)

México City clay MH)

Leonards an d Girault, 1961)

40

+ 20

2-10

6.5-8.7

8.5

0.4-0.7

0.7-3.0

2-4

0.9-1.5

12±6

0.6-3

2.0-2.7

2.7

0.1-0.2

0.2-1.0

0.6-1.2

0.3-0.5

1-1OO

Undisturbed

samples

C v in ra n g o of virgin c o m p re s s i o n

Cy in ranga of

r

«compres

s

¡ en

lies

above this lower l i m i t

Completeiy

remoi e s mples

lies below

this upper limit

4

60 8O 100 120

Liquid limit -

140 160

Figure 5.4 Approximate correlations between

coefficient

of

consolidation

and liquid

limit

after US

Navy, 1988



68 CORRELATIONS

OF

SOIL PROPERTIES

fo r clays a re given in Table 5.6 and an a ppro xima te corre lat ion with

liquid limit is

shown

in

Figure 5.4.

5 3 SECOND ARY COMPRESSION

Secondary compression is a vólume change under load that takes

place

at

constant efíective stress; that

is , after th e

excess pore water

pressure

has

dissipated.

It

is

thought

to

result f rom compression

o f

the co nstituent soil

particles

at a

microscopic

or molecular

scale

and

is particularly

signifícant in

organic soils.

Coefficients of

secondary

compression may

be

defíned

in a way

tha t

is

analogous

to the

definitions o f co mpression Índex and modified compression Índex,

except that the índices are related to time instead of pressure. Thus,

th e

secondary compression índex,

C

a is :

de

~ d ( l o g í )

(5.11)

where

de

is the change in voids ratio o ver a time interval , di f rom time

í

x to time

í2:

see Figure 5.5. Similarly, the modified secondary

compression Índex, C a£ is :

dh/h

d( log í )

(5.12)

o

>

O

su

e

o

e

Primary

con

«oí ¡dat

ion

Secondary compression

Log time t

Figure

5 5

lotting

an d

calculation

of

secondary compre ssion



ONSOLID TION

ND SETTLEMENT 69

where e

p is the

voids ratio

at the

start

of the

linear portion

of the

e-logp or áh — logp curve. The modified secondary compression

Índex

is sometimes also

referred

to as the secondary compression

ratio or the rate of secondary compression.

Calculations o f secondary compression are obtained by rearrang-

in g Equation

5.12:

specimen com pression

dh

becomes secondary

settlment,

p

c

;

specimen thickn ess, h , becomes layer thickness, H; and

the time is taken over a specifíc interval, from í to

í2:

pc

=

CMHlog(t2/í1)

or

For the purpose of secondary settlement calculations, secondary

settlement is assumed to start when primary settlement is substan-

tially complete. Thus, if primary settlements were substantially

complete in 12 years, the valué of

í would be 12. The valué of í

2

depends

on the

assumed lifespan

of the

structure under

consideration.

Valúes of

C

a or CZ£ are obtained from e —

logp

o r

áh— log

p plots,

as indicated in Figure 5.5.

Ca

is usually assumed to be related to Cc,

with

valúes o f

CJCC

typically in the range 0.025-0.006 for inorganic

soils and 0.035-0.085 fo r organic soils. Some typical

valúes

a re given

in Table 5.7. M esri

1973)

obtained a relationship between

C

aE and

natural moisture content, given in Figure

5.6.

Table

5.7

Soil

J

Organic silts

0.035-0.06

Amor phous

and fibrous

peat

0.035-0.085

Canadian muskeg . 0.09-0.10

Leda

clay (Canadá)

0.03-0.06

Post-glacial

Swedish

clay

0.05-0.07

Soft

blue clay Victoria,

B.C.) 0.026

Organic

clays and

silts

0.04-0.06

Sensitive clay, Portland, ME 0.025-0.055

San

Francisco Bay Mud 0.04-0.06

New Liskeard (Canadá) varved clay 0.03-0.06

México City clay

0.03-0.035

Hudson River silt 0.03-0.06

New Haven

organic

clay silt 0.04-0.075

*

M odified

after

Mesri

and Goldlewsk'(197 -



70

CORRELATIONS

OF SOIL

PROPERTIES

10O

ü

M

X

»

TJ

c

o

c o

a

o

u

a

•

c

o

«

o

TJ

•

o

10

1

0 1

10

i i III lili I I I I I I I I I

i i r

r M I

O O

i

f i

i

1

Natural

moisture contení -

Figure 5.6 Correlation between

modified

secondary

compression

índex and

natural

moisture contení

after

Mesri ,

1973)

5 4

S TTL M NT

OF SANOS AND

GRAVELS

5 4 1 Probes and standard penetrador tests

As

mentioned

in the

introductory

re rks to

this chapter

th e

near-impossibility of

obtaining

and

testing

imdisturbed

samples

of

granular soils means that consolidation testing

is not

possible.

Instead settlements are usually estimated from insitu

test

results

most comm only using

th e

standard penetration test althoug h

the use

of

probes

in the form of

static

or

dynamic cones

has

become more




widespread in recent years ESO PT, 1982;

INSITU

1986; ISOPT

1988). A useful review of the interpretaron of some penetration tests

for sands is given by Rober t son and Campanel la

1985).

The most commonly-used correlat ions for set t lement

estímales

in

sands, based

on SPT

results,

are those

established

by

Terzaghi

and

Peck 1967), sho w n

in

Figure 5.7. Terzaghi

and

Peck point

out tha t

the correlations show wide scatter and shou ld not be regarded as

anything more than

a

rough-and-ready guide.

Considering the

practical problems of obtaining me aningfu l SPT results, especially in

sands below the water table, and the disagreements over various

corrections to be applied to the results, th e correlations a re o f dub ious

valué in

many

cases. Yet settlement estimates are of crucial import-

ance for the determination of allowable f ou nda tion p ressures on

granula r

soils,

whose high

u l t ímate

bearing capacity means that

ruu

6OO

CM

E

H

5OO

•

3

S 40O

a

c

OO

0 •

S

200

<

100

—

— —

\

\

•*•».

—

—

e

—

—

ry

den

••^—

— —

se

^

^

s Dense

X

Med

V

íí 30

um d <

S5o

Loóse

.

•••• i»

snse

—

—

—™— —

• -—

• i

••

i.

70

6O

50

4O

«

x

e

c

30 o

4-1

2O

10

O 2 3 4 5 6

Footing width m

Figure 5.7 Chart fo r estimating

allowable bearing pressures

on

sands

using

standard

penetration test results based on 25mm

settlement.

Continuous Unes are based on the

original

chart

by

Terzaghi

an d

Peck

1967);

broken

Unes are iníerpolations




settlement rather

than bearing

failure is the controlling factor. In

view

of all

these considerations it is surprising that settlement calculations

fo r granular

soils have

for so long

relied

on

such

an

unsatisfactory

procedure. Perhaps it

reflects

a

lack

of problems

w ith foundat ions

on

granular soils.

Meyerhof

1956, 1974) also produced relationships between

SPT

results and

settlement which gave similar valúes

to

those

o f

Figure

5.7. However, both

th e

Meyerhof

and the

Terzaghi

and

Peck valúes

are considered to be conservative, and Bowles 1982) suggests that, in

th e

light

of field

observations

and the

stated opinions

o f many

authors, th e Meyerhof equations should be

adjusted

to

give

an

approximate

50 increase in allowable

bearing

capacity for 25mm of

settlement qa), thus:

for foundation w id ths

metres,

up

to

1.2m

05 d

4

a

(kN/m

2

)

=

N

fo r

foundation widths

metres,

greater

than

1.2m

.08

where N

is the SPT N-valué standard blows per 300mm)

K

á

=

1 +0.33D/5 up to a máximum valué of 1.33

and D is the

depth

to the

foundation base,

in

metres.

Plots of

these equations,

for

D

= 0

i.e.

a surface

foundation)

are

shown

in

Figure 5.8.

For

founding

depths

up to

D

=

B

valúes

obtained from this chart

may be

multiplied

by

K

á

. Terzaghi

and

Peck

suggest that,

fo r

saturated sands, allowable bearing pressures

ob-

tained from Figure 5.7 should be reduced by a half for shallow

foundations

and by a

third where depth

D is

approximately equal

to

width B.

Bowles 1982) gives

no

mention

of

such reductions

but it

seems

prudent to also

apply

them when using th e above equations

and Figure 5.8. Allowable bearing

pressures

for settlements other

than 25mm

may be

obtained pro-rata.

Raft

foundations are known to settle

less than strip

footings, and

Tomlinson 1980) suggests that

the

allowable settlements obtained

from Figure 5.7 be doubled for this type o f foundation. Alternatively,

Bowles gives

a mo dified

form

of the

Meyerhof equation

for rafts:

N

Work

by

Menzenbach 1967) established

a

rough relationship

between deformation modulus, E¿ and SPT

N-value,

as shown in

Figure 5.9. This can be used in conjunction with

elasticity

theory to



CONSOLIDATION AND

SETTLEMENT

73

8

1 2 3 4

Footirvg width

m

Figure 5.8

Allowable

bearing pressure fo r

footings

founded ai surface level, for

settlement limited

lo approximately

25mm

after

Bowles, 1982)

obtain sett lem ent

predictions. For

instance

for a strip

foundation

o f

width B,

loading

intensity

q

settlement p

is

given

by:

2 25

where

Poisson s

ratio v is usu ally taken as 0.15 for sands. Valúes of



74 CORRELATIONS

OF SOIL

PROPERTIES

8

6

•o

O

40

2

l

Overburden pr ssur kPa

2O 40

SPT

N value b l o w s / S O O m m

60

Figure 5.9

Correlation

between

deformation

modulas,

E

d and SPT

N-value

for granular

soils after

Menzenbach, 1967

allowable bearing pressure

fo r

25mm settlement obtained

in

this

way

a re

broadly

in

line

with

the

valúes obtained

from

Figure 5.8.

It should be noted that although the rate of settlement is not

determined from SPT results the high permeability of granular soils

produces rapid response to loading so

seti-

ment times a re very short

a nd rarely considered.

5 4 2 Píate bearing

t sts

Píate bearing tests offer a more direct method of measuring settle

ments but the usefulness of the results is limited by tw o constraints:



CONSOLIDATION

AND SETTLEMENT 75

1) the depth of sand stressed by a píate is

only

a fraction of that

stressed

by a

full-sized foundation,

and

2) settlement predictions require knowledge of the scale effects

between

the settlement of a píate and that of a full-sized

foundat ion.

The

most

commonly-used

correlation

fo r

scale

effects

between

píate

and

fou ndation settlements

is

that given

by

Terzaghi

and

Peck

1967):

where

p is the settlement of a square foundation o f side B ft, and

pt

is the settlement of a 1-foot square píate

If th e foundat ion width is measured in metres, this becomes:

2B

0 3

A n alternative, and more general, relationship was derived by

Menard and Rousseau 1962):

P i =

P2

where pí

and

p

2 are the

settlements

of the

píate

and

footing

B±

and

B

are

the ir respective w idths

and a depends on the soil type. Ty pical a valúes are:

Sands and gravéis to ^

Saturated

silts

Clays and dry

silts

§ to

Compacted fu l 1.



hapter

SHE R STRENGTH

It is usually assumed that the shear stren gth of soils is governed by the

Mohr-Coulomb

failure

criterion:

s = c +

< 7

tan >

6.1)

where s is the shear stress i failure along

any

plañe

a is the normal stress on that plañe

and c and f ) are the shear strength parameters; cohesión and

angle

of

shearing resistance.

This is sho wn graphically on the

Morir

diagram given in Figure 6.1.

A complication arises because th e norm al stresses within a soil are

c rried

pa rtly b y the soil skeleton itself and p artly by water within th e

soil voids. Considering only th e stresses

within

th e soil skeleton,

equation

1)

is

modifíed

to

or

s = c

+

a

tan

>

where u is the pore water pressure

a

=

(a—u),

th e

effective norm l stress

on the

soil skeleton)

and c an d < / > are the shear strength parameters related to effective

stresses.

Th us wh en considering the shear strength of soils, ther e is a cho ice:

either th e total, combined reponse of the soil and pore rater can be

considered Eq uation 6.1);

or the specific

response

o f the

s « l skeleton

can be separated from the pore water pressure by

considen - . effective

stresses Eq uation 6.2).

The

effective stress appro ach gives

a

truc

measure

of the

response

of

th e soil skeleton to the loads imposed on it. Perhaps th e

simplest

case

is that of a load applied to a saturated soil that is allowed to drain. If

the rate of application of the

lo d

is

sufficiently

slow,

pore water

76



S H E A R

STRENGTH

77

Figure

6.1 Mohr diagram representing th e

g eneral Mohr-Coulomb failure

criterion

ire t

str ss

Figure 6.2 Mohr diagram for a normally-cons olidated clay for effective s tresses

pressures will

not

bu ilt

up and the

tota l stresses will equ al

the

effective

stresses Fo r drained conditions, or in terms of effective stresses, it is

found

that the shear strength of

soils

is principally a frictional

phenomenon, with

c = 0, as

ülustrated

in

Figure 6.2. This does

not

appear to b e the case f or ov erconsolidated clays which have a

bu ilt-in

pre-stress

(see Singh et al. 1973),

or for

partially saturated clays

in

which

th e particles are

drawn together

by

surface

tensión

effects,

giving them some cohesión.

When soil

is

loaded,

th e

increase

in

confming pressure within

th e

soil skeleton squeezes the particles closer together, reducing the

volume of the voids. However, in a saturated clay this cannot take

place unless some of the pore water can drain f rom the voids.

Thus ,

for

a saturated clay in c onditions of no d rainage, an increase in

confining

pressure cannot be carried by the soil skeleton but results

instead in an equal increase in pore water pressure. Since shear

strength depends on the effectiv e stresses, trans m itted by interparticle

contacts, and

these

remain unchanged

irrespective

of the applied

confining pressure, it follows that undrained shear strength will also

be independent of confining pressure. Because of this, samples of

saturated clay tested

in a

quick undrained triaxial test

give Mohr s

circles

of co nstant diameter and an apparent c ohesión v alué as

shown



78 C O R R E L A T IO N S

OF

SOIL PROPERTIES

xjjl---Effective

s t r e s s failure

e n v e lo p e

Total stress f i lu re enve lope

Figure 6.3 Mohr diagram for s tur ted clay in terms of tot l an d effective síresses

in Figure 6.3, even though, in effective stress terms, the material is

basically

frictional.

Thus ,

in a sense, the phenomenon of cohesión is

an illusion brought abouí

by

the response of pore

water

pressures to

imposed

loads.

T o

underline

this poin t, the term apparent cohesión

is often

used. Partially saturated soils, tested in undrained co nditions,

will show a behaviour whic h is intermedíate between that for drained

co nditions and f or saturated und rained co nditions, depending on the

degree of

saturation.

6 1

THE CHOICE OF TOTAL OR EFFECTIVE STRESS

ANALYSIS

When the

soil

is

loaded rapidly

so

that there

is no

t ime

for

movement

of pore w ater to tak e place, its imm ediate response - the proport ions

of

the resulting confining pressures that are carried by the soil

skeleton and the pore water - is itself a property of the soil. This

instantaneous response can,

in

fací ,

be quantifíed in

terms

of

Skempton s 1954) pore pressure parameters, which are described in.

Chapter 5 . This m eans tha t the total response of the soil to an applied

load, including

th e

pore pressures generated,

can be

s imulated

and

measured

in a

laboratory test

a nd

there

is no

need

to

take account

o f

the

sepárate responses of the

skeleton

and the pore

w ater.

Only the

total applied stresses need be considered in the analysis and only the

corresponding total stress strength parameter~ need be measured

when testing.

Strictly

speaking, this

is not

qui

;

true because soil

strength is usually measured in the triaxial

test,

in which axially

symmetric

stress conditions exist, whereas many soil problems

approximate to plañe strain conditions, for which the soil response

diíiers slightly,

but the

errors involv ed

are

small enough

to be

ignored

for practical purposes.



SHEAR STRENGTH 79

The equilibrium

pore water pressures that

are

eventually estab-

lished

are,

unlike

the imm ed iate response, not a pro perty of the soil

but depend on the surrounding conditions. Long-term pore water

pressures cannot therefore be simulated in the laboratory must be

considered separately. Henee, efíective stress analysis must

be

used

where

long -term

stability

is

important .

In

testing,

the

response

of the

soil skeleton can be

measured

either by allowing drainage of the

specimen so tha t no m ore pressures build up or by measuring the pore

water pressure within the specimen. In either case, tests must be

carried

out

slowly enoug h

to

allow c omp lete dissipation

or

equalisa-

tion

of

excess

pore

water pressures within

the

test specimen.

6 1 1

he

choice

in

practice

Foundations impose both shear stresses

and

compressive stresses

confining pressures) on the und erlyin g soil. The shear stresses m ust

be

carried

by the

soil skele ton

but the

com pressive stresses

are

initially

carried largely by the resulting increase in pore w ate r pressures. This

leaves

th e effective

stresses little changed, which implies that

th e

foundat ion

loading

is not

accompanied

by any

increase

in

shear

streng th. As the excess pore pres sures d issipate, the soil

consolídales,

and

effective

stresses increase, leading to an increase in shear str en gth .

Thus,

for

foundations,

it is the

short term co ndition

the

imm ediate

response of the

soil

that is mo st critical.

This

is the justifícation for

the use of quick undrained shear strength tests and

total

stress

analysis

for

foundation design.

W ith excavations, com pressive stresses are reduced by removal of

soil but shear stresses are imposed on the sides of the exca vation

owing

to removal of lateral

support.

Initially, th e reduction in

compressive stresses

is

manifested within

the

soil mainly

as a

reduction

in

pore water pressures, with little change

in

eífective

stresses so th at , as with foun da tion s, soil shear strength remains little

aífected

by the changed loading. Ev entua lly, wa ter flows into the soil

that

forms

the excavation sides, restoring th e

pore-water

pressures.

This reduces the

effective

stresses, causes swelling and reduces shear

strength. Thus,

for

excavations, long-term conditions

are the

most

critical. Since long-term pore pressures depend

on

drainage condi-

t ions and can not be simulated by soil tests, an eífective stress analysis

must be used so that pore water pressures can be

considered

separately from stresses

in the

oil skeleton.

During embankment construction, additional layers of material

impose a pressure on the

lower

part of the embankment. As with

foundations, this tends to créate increased pore water pressures and,



80

CORRELATIONS

OF SOIL

PROPERTIES

by

the same argument, short-term conditions are an

important

consideration. This implies that total stress analysis and quick

undrained shear strength tests are approp riate, and

up

to the 1960s it

w as not

uncommon

for

emb ankments

to be

designed

in

this way.

However,

additional stresses

can be

created

by the

compaction

process itself but,

offsetting

this,

th e

material

is

unlikely

to be

saturated

so

that

a significant

proportion

of the

added pressures

m ay

be carried

immediately

by the soil

skeleton. These complications

make it impossible to simúlate the total response of the soil in a test

specimen and, to overeóme this,

effective

stress analysis is now used.

Also it is usually more economical to design embankments fo r

long-term stability

and to

monitor pore water pressures during

construction, slowing dow n the rate of construction

where

necessary,

to

keep

them

within

safe

limits.

A

special case

o f

emban km ent stabili ty,

often

quoted

in

text books,

is that of the rapid drawdown of water

level

behind an embankment

dam. In this case, the soil in the embankment has had time to

consolídate

under

its ow n

weight implying long-term cond itions)

but

support from

th e

adjacent water

is

w ithdraw n rapidly implying

short-term conditions). This can be simulated by the C onsolidated

undrained triaxial test,

in

which

th e

test specimens

are

allowed

to

drain and

consolídate

under the applied

cell

pressure. Once consoli-

dation

is

complete, specimens

are sheared

rapidly un der conditions

of

no drainage. In this way, the response of the soil to both long-term

consolidation

and

short-term shearing

is

simulated

in the

test,

allowing a

total

stress analysis to be used. The simulation of

long-term conditions

in a

test

is

assumed

to be

possible

in

this case

because the water in the reservoir ensures that the soil on the

up-stream

face

of the dam will al w ay s be saturated. However, the

rapid drawdown condition can be better more thoroug hly, analysed

in terms of eífective

stresses, using

the

effective

stress strength

parameters which musí be measured anyway for n ormal long-term

stability analysis

of the dam

slopes.

The use of the

Consolidated

undrained test witho ut pore pressure measurem ent is therefore more

of

historical

interest than practical application.

With

natural slopes,

we are

alw ays dealing

with

conditions that

have

been in equilibrium for a long period of time, although seasonal

variations will occur,

an d effective

stress analysis

is

appropri

e.

6 2 UNDR INED

SHEAR STRENGTH

OF

CLAYS

Shear strength is obtained

from

the

M ohr-Coulomb failure criterion,



S H E A R S TR E N G TH

81

Table 6 1 E S T I M A T I N G THE S H E A R S T R E N G T H O F C L A Y S

Shear

strength

kN / m

2

)

<20

20-^W)

40-75

75-150

150-300

>300

Descriptive

term

Very so f t

Soft

Firm

Stiff

Very stif

Hard

Characteristics

Exudes between f ingers when squeezed

Moul ded

by

light

finger

pressure

Moul ded by s t r ong f inger pressure

Can be indented by t h u m b

Can be

indeníed

by

t h u m b nail

Note: thesc

strength

descriptions and tests conform w i th standard practice and with the recommendations of B.S.

5930 1981).

Table 6 2 T Y P I C A L S H E A R S T R E N G T H P R O P E R T IE S O F C O M P A C T E D C L A Y S

Soil

description

Class

Undrained shear

strength

kN/m)

As

compacted

Saturated

Silty

sands sand-si l t m ix

Clayey sands sand-clay m ix

Silts and clayey silts

Clays o f

lo w

plastici ty

Clayey silts, elastic silts

Clay of high plasticity

SM

SC

M L

CL

M H

CH

50

74

67

86

72

103

20

11

9

13

20

11

Uniíied

classif icat ion system.

Equat ion 6.1). Ho wever, fo r mos t sa turated

clays,

tested under quick

undrained condi t ions ,

the

angle

o f

she aring resistance

is

zero.

This

means that the shear s t rength of the

clay

is a fixed valué and is equal to

the apparent coh esión.

T he

valué

o f the

undrained shear s t rength

may

be

est imated

by

mould ing

a

piece

o f

clay between

the fingers and

applying

the observations indicated in Table 6.1.

Typical

valúes

for the

shear

s t rengths of

compacted

clays

are

given

in Table 6.2. Valúes

refer

to soils compacted to the m á x i m um dry

densi ty obta ined in the s tandard compact ion tes t : AASHTO T99

5.51b

rammer method)

or BS

1377:1975 Test

12

2.5kg ramme r

method) .

6 2 1 Remoulded shear strength

As discussed

in

Chapter

1, the

liquid

and

plástic limits

are

mois ture

contents

at

w hich soi l

has

specific valúes

o f

undrained sh ear s t rength.



82 CORRELATIONS

OF

SOIL PROPERTIES

2

1.8

1 6

1.4

x 1.2

o

1 0

2

cr

0 8

6

4

2

Liquid

limt

0 .2

Plástic

l i m i t

Clay

Horten

London

Gosport

Shellhaven

LL

PL

30

16

73

25

80 30

97 32

Pl

Activity

14 0 36

48

0 96

50 0 89

65

1 27

I I I l i l i

J

I I I I I

l i l i

1

O.5 1 5 1O

Undrained

shear strength kN/ra

5O 100

2

Figure

6 .4

Correlation between shear strength

and liquidity índex after Skempton and

Noríhey,

1952)

It therefore follows that for a rem oulded soil the shear strength

depends on the valué of the

natural

m oisture

c ontení

in relation to the

liquid an d

plástic

limit valúes. This can be convenie ntly expressed by

using

th e

concept

of

liquidity índex defined by :

w

n

PL wn PL

Liquidity índex = -—

P

Pl



SHEAR

STRENGTH 83

where

LL and PL are the liquid and

plástic lim its, respectively

PI is the

plasticity

índex

and w

n

is the natural moisture content.

Curves relating remoulded undrained shear strength to liquidity

índex

hav e been established by Ske m pton and No rthe y 1952). These

are given in Figure 6.4.

6 2 2 Undisturbed shear strength

The shear strength o f undisturbed clays depends on the consolidation

history of the clay as well as the fabric characteristics.

The ratio of natural shear strength to remoulded shear strength is

known

as the

sensitivity.

It is

most marked

in soft,

lightly con-

solidated clays which have

an

open structure

and a high

moisture

content

Sensitivity m ay

be

related

to

liquidity Índex,

and

this

has

indeed been found so by a num ber of researchers, whose findings are

given and discussed by Holtz and K ovacs 1981). M uch of this data is

for the

sensitive clays

of

Canadá

and

Scandinavia

but the

work

of

Skempton and Northe y 1952) relates m ainly to clays of relatively

modérate sensitivity with natural mo isture co ntents

below

th e liquid

limit. Their fíndings are

given

in

Figure 6.5.

Fu rther, since both remoulded shear stren gth

an d

se nsitivity

can be

correlated with liquidity Índex, it foliows that a

correlation

m ust exist

between undisturbed shear strength and liquidity índex. Such a

relationship,

obtained

by

combining

the

correlations

given

in

Figures

2

5O

20

§ 1

0 2 O 0 2 0 4 0 6 O 8 1 0 1 2 1 4 1 6 1 8 2 0

Liquidity ndax

Figure 6.5 Correlation between sensitivity and liquidity

índex

after Skempton an d

Northey, 1952)



84

C O R R E L A T I O N S OF SOIL PROPERTIES

200

100

g 5O

x

JS

O

e

o

o

1

i

I I I 1

.

ti

0 2 0 4 0 6 0 8

iquidi ty indax

1

1 2

2

a

Figure 6.6 Relationship

between

th e natural shear slrength of undisturbed clays and

liquidity índex

6.4

and

6.5

is

shown

in

Figure 6.6 which then provides

a useful

predictive tool for assessing the shear strength of un disturbed soils.

It is

found

that for

m ost norma lly-consolidated clays, und rained

shear strength is

proportional

to eífective overburden pressure. This



S H E A R STRENGTH 85

is to be

expected when

it is

remembered that ,

in terms of

eñective

stress,

shear

strength is basically a

frictional

phenomenon and

depends on confming pressure. If

th e

constant of proportionality

between shear strength

and

eñective overburden pressure

is

k n own

then shear strength

can be

inferred

from

eñective overburdenpressure; that is ,

from

depth.

This

problem h as been investigated by a

number

of

researchers, with

a view to

establishing

a

correlation

between

the

shear strength/ove rburden pressure ratio

and

some soil

classification param eter, typically the plasticity índex. Such a correla-

tion would be of great practical valué, since it would enable the

undrained shear strength (S u)

to be

estimated

from a

simple

classification test.

Historically,

muc h use has been m ade for normally consolidated

clays of the relationship of Skempton 1957):

<7V

0.11 0.0037P/

where,

P I

is the plasticity Índex. At first sight it is not evident that

SJ(j v should b e related to the plasticity Índex. Ho we ver, the valué of

0

can be

expected

to

depend

on the

shape, size, packing

and

mineral

composition of the clay particles, as

will

the plasticity

índex,

so the

two properties

are

related

in

some man ner see Figure 6.12). Figure

8

i

0)

n

Bjer rum(1972 ) aged

Skempton

(1957)

Bjerrum (1972) young'

Kenn0y(1976)

100

Plas ticity index

200

Figu re 6.7 Relationship betwee n the ratio of undrained shear strength to effective

overbu rden pressure

an d

plasticity index

for

normally-consolidated clays

(modified

after

Holtz and

Kovacs ,

1981).



86


OF

SOIL P R O P E R T I E S

6.7 includes

other results obtained

by a numb er of

researchers.

As can

be seen, their findings vary and should only be used wi th caution.

Ho w ever, such correlations particularly that

of

Skempton 1957)

are

useful for preliminary estímales and checking laboratory data on

normally Consolidated clays. For overconsolidated

clays,

Kenney

1959), stated that the relationship is influenced mainly by the stress

history and is essentially independent of plasticity Índex. A correla-

tion between

the

shear strength/overburden

pressure ratio and

liquidity

índex for

Norwegian quick clays

w as

presented

by

Bjerrum

and Simons 1960), as

indicated

in Figure 6.8. Ag ain, results show so

much

scatter that the interp retatio n of the results is open to q uestion,

and all that can be said with certainty is that, for Norwegian quick

clays, the

ratio

is around 0.1 to 0.15.

Besides

th e

influence

of geological history on undrained shear

strength,

the

stress history du ring test also

affects

results. Thus, shear

strengths obtained

by

unconfined

compression testing or triaxial

testing can be expected to difíer from those

obtained

by shear vane

Wroth, 1984). The relative valúes of the

shear

strengths have been

examined by a number of researchers, and the

ratio

of

true

und rained shear strength based on the back-analysis of em ban km ent

failures)

to shear vane valúes seems to depend on the plasticity índex,

as indicated by Figure 6.9.

Strictly, undraine d shear strength depends on the

effective

consoli-

dation pressure, which is the average of the

effective

overburden

o »

o

O

»

o

0 4

0

°3

•o

0.2

w

o

Liquidity Ín ex

Figure 6.8

Relationship between

the

ratio

of

undrained

shear

strength

an d effective

overburden

pressure and liquidity Índex for Norwegian clays after Bjerrum and

Simons, 1960)



S H E A R STRENGTH 87

3k

o

v.

u.

Ü

1.4

1.2

1.0

0.8

6

0.4

D O

•

Bjerrum 1972)

Milligan 1972)

Ladd and

Foott 1974)

-

Flaate

and Preber

1974)

LaRochella et al. 1974)

D Holtz and Holm

1979)

*

- Layered and varved clays

B j e r r um s 1972)

recommended curve

« CH

20 40 6O

P l a s t i c i t y

í n d e x

80

1 0 O

120

Figure

6.9

Correlation factor

fo r

field vane test results, depending

on

plasticiíy índex ,

basedon b ack-analysis of embankment failures after Ladd, 1975 and Laddet al. 1977)

pressure

and the lateral pressures. For overconsolidated clays,

comparison of shear strength with effective consolidation pressure

gives better correlations than with

effective

overburden pressure.

According to Bjerrum (1972), w ork ing with normally-consolidated

late

g lacial clays , w hilst recent sediments

are normally

Consolidated,

older clays tend to be slightly overconsolidated, the overconsolida-

tion ratio depending somew hat on the plasticity Índex, as indicated in

Figure 6.10. Combining this w ith Bjerrum s shear

strength/overbur-

den

pressure relationships (Figure 6.7),

and

correcting

th e

resulting

shear strengths using the factor

//

from

Figure

6.9

Mesri (1975)

concluded

that the ratio of the field shear strength to effective

consolidation pressure was independent of plasticity índex and was

equal to 0.22. The scatter of results which ha

ve

gone into producing

this conclusión

are so

w ide that

it

mus t

be

viewed with great caution

but, if validated, it

could

be of practical valué.

Although the literature contains m uch debate

concerning

S

u

/a;

and

overconsolidation ratios

(Ladd

e t al,

1977; W roth , 1984), in p ractical

terms it is more straightforward to measure the undrained shear



88

CO RRE L A T I O N S OF SOIL PRO PERT IES

2

2

40 6O

Plasticity índex

1

Figure 6.10

Relationship between overconsoliation ratio an d plasticity

Índex

fo r

late-glacial clays after

Bjerrum,

1972

5

4

3OO

•H

20 0

0

•a

D

.

Soil

groups refer

to

Unified

system

Terzaghi and Peck

1O 2 3O 4 5

SPT N valué blows/SOOmm

6O

Figure 6.11 Approximate correlations beíween undrained shear strength and standard

penetration test N-values

after

Terzaghi

an d Peck,

1967

and

Sowers,

1979}

strength of overconsolidated clays tha n to predict it from other

índices.

6 2 3 Predictions using the standard penetration test

Attempts

have been made

to

correlate

th e

unconfined compressive

strength

or the

undrained shear strength

of

clays with

the

results

of

standard penetration tests with varying degrees of success. Some

suggested relationships are

given

in Figure

6 11



S H E A R

STRENGTH 89

DR INED AND EFFECTIVE SHE R STRENGTH OF

CL YS

As discussed previously it is often impor tan t to carry out stability

calculations in

terms

of

effective

stresses. This is particularly truc o f

slope

s tabili ty calculation s.

T he

soil streng th param eters

used in

these

calculations are obtained f rom ei ther drained shear box or triaxial

tests (giving cd and < / > d or

f rom

Con solidated un drain ed triaxial tests

with

pore pressure measurement (giving < / > é

u and c

cu

). I n

theo ry there

should

be little

difíerence

between the tw o sets o f valúes, for sa turated

clays, although in practice there

may

be

minor differences.

*

A

relationsh ip between

diaÍnecLshjea£.stEejftgth and

p lasticity Índex

for remoulded clays

has

been established

by

Gibson (1953),

as

indicated in Figure 6.12. Also shown is a relationship between th e

residual shear strength, or true angle of internal

f r iction,

an dplasticity índex. The existence of these relationships arises because

both plasticity Índex and shear strength reflect the clay mineral

composition of the soil: as the clay mineral content increases,

4O

3

£ 2

w

o 1

o

i

Drained »h««r ¿

d

[_U ^M

Truo

angl©

of

internal friction

i

/

2

4

6 8

Plasticity indox

1

Figure

6.12 Relationships between angle

of

shearing resistance

and

plasticity Índex

after Gibson, 1953)

12



90 C O R R E L A T IO N S OF

SOIL

PROPERTIES

Table 6 3

TYPICAL

A N G L E S OF E F F E C T I V E S H E A R I N G R E S I S T A N C E F OR C O M P A C T E D C L A Y S

Soil description

Class*

d e g )

Silty clays,

sand-silt

m ix

Clayey

sands,

sand-clay m ix

Silts and

clayey

silts

Clays

of

lo w

p lasticity

Clayey silts, elastic silts

Clays of

high plasticiíy

SM

SC

M L

CL

M H

CH

32

28

25

19

* Unified classification system.

plasticity índex mercases and shear strength decreases. As described

previously, the strength of clays, in eñective stress

terms,

is basically

frictional so

c =

0.

This

is

certainly

th e

case w ith rem oulded saturated

clays but

partially saturated clays, where meniscus

effects

draw

the

particles

together to produce inter-particle stresses, m ay appear to

have a

small

cohesión valué, though this itelf is a frictional

phenomenon.

Typical valúes of the angle of shearing resistance, 0 , for compacted

clays are

given

in

Table 6.3. Valúes

are for soils

compacted

to the

máx imum dry density according to the standard compaction test

(AASHTOT99,5.51brammermethod;orBS

1377:1975

test

12,2.5kg

rammer

method) .

6 4 SHEAR STRENGTH OF GR AN UL AR SOILS

Because

o f

their high perme ability, pore w ater pressures

do not build

up when granular soils are subjected to shearing forces, as

they

do

with clays. The com plicátion of total and effective stresses is therefore

avoided and the pheno m enon of apparent cohesión, or undrained

shear strength, does

no t

occur. Consequently,

the

she ar strength

o f

granular soils is defíned exclusively in terms of the frictional resistance

between the grains, as measured by the angle of shearing resistance.

Typical valúes

of the

angle

of

shearing resistance

fo r

sands

and

gravéis are given in

Table

6.4.

Typical valúes

for

compacted soils

are

given

in

Table 6.5. Valúes

refer to soil

com pacted

to

m á x i m um

dry

density

at

opt im um mois ture

content

as defíned in the standard compaction test: AASHTO T99

5.51b

rammer method)

or BS

1377:1975 test

12

2.5kg ramm er

method) .

A


dry

density

or

relative de nsity

and the

angle

of shearing resistance is given by the US Navy 1982) , as show n in



S H E A R

STRENGTH 9 1

ble

6 4

T Y P I C L V L Ú E S OF THE N G L E O F S H E R I N G R E S IS T N C E O F C O H E S IO N L E S S

S O I L S

Material

Unifo rm sand, round grains

Well-graded sand, angular grains

Sandy gravéis

Silty

sand

Inorganic

silt

Loóse

27

33

35

27-33

27-30

deg)

Dense

34

45

50

30-34

30-35

ble 6 5 TYPICAL V L Ú E S OF THE N G L E O F S H E R I N G R E S I S T N C E F O R C O M P C T E D

S ANDS

AND

G R A V E L S

So// description Class*

Angle

of

shearing

resistance, f > deg)

Well-graded sand-gravel mix tures

Poorly-graded sand gravel m ixtures

Silty gravéis, poo rly graded sand-gravel-silt

C layey gravéis, poorly graded san d-gravel-clay

Well-graded clean sand gravelly sands

Poorly-graded clean sands, gravelly sands

GW

GP

GM

GC

SW

SP

>38

>37

>34

>31

38

37

1 Unified classification

system.

O

O

c

o

50

O

• 40

a

c o

30

20

Material

type Unified classification)

Relative

density

1.2

1 4 1 6 1 8 2 0

Dry

density

-

t/m

3

Mg/m

3

)

2.2 2 4

Figure 6.13 Typical valúes ofdensüy an d angle o f shearing resistance o f cohesionless

soils modified after

US

Navy 1982)



92


o

3

a

8

5

4

2

10

4

X

X

x

x

X

/

/

Relative density

Very

dense

Loóse

.

t*

ery

loóse A,

28 3 32 34 36

38

4 42

44 46

g of

shearing resistance °

Figure 6.14 Estimation

of the

angle

of

shearing resistance

of

granular soils from

standard

penetration test result after Peck

et

ai 1974)

Figure 6.13.

The

material types

indicated in the figure

relate

to the

Unified

classification

system . Peck et al. 1974) give

a

correla tion with

standard penetration test valúes, shown

in

Figure 6.14.

The

correla-

tion between

SPT

valúes

and

relative den sity

is

also

shown , enabling

a

comparison

to be

made with

the U S

Navy valúes.

Examination

of

Figures 6.13

and

6.14 show s reasonab le agreem ent

between the two correlations. H ow ever, considerable variation can

exist within each soil type, as indicated by Figure 6.15, which shows

plots

o f the

angle

of

shearing resistance against relative density

for a

number of sands.

6 5 L TER L PRESSURES IN A SOIL

MASS

Consideration

of

lateral pressures

is usu;-lly

associated with

the

design

of

retaining walls, basement walls pile foun datio ns

and

tunnels, where interest is centered on the m¿ iñ m um nd máximum

lateral

pressures

that

can

occur;

that is, on the

coefficients

of active

and passive pressure. Approximate solutions for active and

passive

pressure problem s can be obtained using the simple Cou lomb 1773)

wedge theory or by consideration of Mohr s circles of stress at failure

Rankine, 1857). The R ankine

approach

is still used for cohesive and



S H E A R S T R E N G T H

93

<

e

o

O

O

e

20

O

60

Relativo density -

Figure 6.15 Relationship s beíween angle ofshearing resistance and

relaíive

density for

various sands after

Hilf,

1975)

cohesive granular c — < / > soils but both the Rankine and Coulomb

methods give

signifícant

over-estimates

of

lateral pressure

for the

passive condition and for granular soils, i t is more usual to obtain

coefficients

of earth pressure using analyses that postúlate curved

failure surfaces Caq uot and

Kerisel,

1966; Terzaghi and Peck, 1967).




0.8

o

0 6

O

o

t_

O.4

o

O.2

U

D

Sangamon sand subang u la r )

•

W a b a s h s a n d s u b a n g u l a r )

O

Cha ta hooche e sand subangu la r )

Bras ted sand

o

Sand

Simons, 1958)

Belgium sand

4- Minnesota sand rounded)

X

Penn sy lvan ia sand angu la r )

O

28 30 32 34 36 38 40 42 44

Angle of shearing resistance, 0 - degrees

46

Figure 6.16

Correlation between th e coefficient of earth pressure at rest and the

angle of shearing resistance for normally-consolidated sands after Al-Hussaini and

Townsend,

1975}

0.8

K

n = 1 -

sin0 ±0.5

0.3

12 14

Ang le

of

s h e a r in g r e s i s t a n c e , 0 - degrees

Figure 6.17 Corre lation beíween the coefficient of earth pressure at rest and the angle

of shearing

resistance,

in

terms

ofeffective

stresses

after

Laddet al. 1977).

Key ío

data:

1) Brooker an d Ireland 1965), 2) Ladd 1965), 3) Bishop 1958), 4) Simons 1958),

5)

Campanella

and Vaid

1972),

6)

Compiled

by

Wroth 1972),

7)

Abdelhamid

an d

Krizek

1976)



SHEAR STRENGTH 95

1.0

O

••

0.8

O

a

§ 0 6

m

0.4

o

ó 0.2

o

K

0

=

0.44 0.42 PI/100)

o

o o

•

Undisturbed

Disturbed or laboratory reconsolidated

from a

sediment

20 40 60 80

Plasticiiy

índex, Pl

100 120

Figure 6.18 Correlaíion between the coefficient ofearthpressure ai rest -

obtainedfrom

laboratory

tests and plasücily índex afíer Massarsch,

1979}

Active

and passive

pressures

represent the limit ing valúes of

lateral

earth pressure, w h e n

the

soil

has reached a

failure condi t ion ,

and

require a certain a mo unt of mo vem ent for pressures to at tain these

valúe s. This

can be of

practical impo rtance

in the

calculat ion

of

design

pressures behind rigid structures, such as strutted retaining

walls,

in

which

m ovem ent m ay

be

insufiícient

to

allow

the

soil

to reach a

passive state.

F or

such condit ions,

it is

useful

to be

able

to

est imate

t he

valué of hor izon tal s t ress in the un disturbed ground . This can not be

obtained from theoret ical considerat ions of limit equilibrium, as is

th e case fo r active and passive

pressures,

but

depends

on the

geological history o f the soil . H ow ev er,

using

an

approx imate

theory

Késdi , 1974

th e

coefficient

of

earth pressure

at

rest ,

K

Q

for a

normally-consolidated soil

can be

related

to the angle of

shearing

resistance:

This relat ionship has been found to

hold

t rue for normally-con-

solidated sands and clays, as indicated in Figures 6.16 and

6.17.

In

addi t ion, a relat ionship between K0 and plasticity

índex

has been



96 C O R R E L A T I O N S O F

SOIL

P R O P E R T I E S

3

2 8

2 6

2 4

2 2.2

o

O

o

2.0

3

£ 1.8

a

£

C

9

°1 4

~ 1 2

Ü

6

0 4

i

T í

o Boston blue

clay, Pl=23

Ladd, 1965)

Brooker

and

ireland 1965)

Plasticity

índex

s

34 6 8 10

Overconsolidation ratio

2O 3O

Figure 6.19 Correlation between c oefficient

of

earth

pressure at

rest

and

overc onsolida-

tion ratio for clays of various plasicity índices data by Ladd, 1965, and Brooker and

Ireland, 1965; replotted by Ladd, 1971

obtained by M assarsch 1979), as shown in Figure

6.18.

The above

relationships are valid for normally Consolidated clays but for

overconsolidated clays the valué of

KQ

is heavily depend ent on the

overconsolidation

ratio. For

these clays,

K0 can be

estimated from

Figure

6.19

wh ich shows relationships between K

0 and

overconsoli-

dation

ratio

fo r clays of different plasticity

índex

valúes.



Chapter

7

CALIFORNIA BEARING RATIO

7 1

TH

T ST

METHOD

The CBR

test

was

originally

developed at the

California División

of

Highways in the 1930s as parí of a study of pavement failures. Its

purpose was to provide an assessme nt of the relative stability of fine

crushed

rock

base

materials.

Later

its use was

extended

to

subgrades.

It

is now

widely used

for

pavement design

throughout the

world.

Ironically it was used for pavem ent design in California for only a

few

years

and was

superseded

by the

Hveem Stabilometer test.

During testing

a

plunger

is

made

to

penétrate

th e

soil which

is

contained

in a

standard

mould at a specified rate of

penetration.

The

resulting

load-deflection

curve

i s

compared

with

that obtained

for a

standard crushed rock. The test details ha ve been largely standar-

dized

and are

given

in the

AASHTO Standard Speciíications Test

T193

and in BS

1377:1975 Test

16 .

Slight variations exist between

th e

Am erican

and

British standards

but

these should have little

effect

on the CBR

valúes

and

arise purely

as a

result

of

converting

the

U.S.

specifícation to metric

units. However significant variations

in

sample preparation and test procedures can occur even within the

specifications. Th is can give rise to difficulties wh en comparing CBR

results from different sources. Table 7.1 show s some of the variations

between

methods.

The CBR

test

is

used exclusively

in

conjunction with pavement

design methods

and the

method

of

sample preparation

and

testing

must

relate

to the assumptions made in the design method as well as

to

assumed site conditions.

For

instance

th e

design method

m ay

assume that soaked CB R valúes are alw ays used regardless of actual

site conditions.

7.2

CO RRELATIONS WITH SOIL CLASSIFICATION

SYSTEMS

In view of the fact

that

early

pavem ent design methods were based

on

soil classification tests rath er trian

CBR

valúes

it

seems

a

reasonable

97



98 C O R R E L A T I O N S OF SOIL P R O P E R T I E S

ble 7.1 V R I T I O N S OF T E S T

M E T H O D

FOR CBR T E S T

Density

The CBR is usually

quoted

for the assumed density of the soil in place. This

will

typically

be 90 , 95 or 100 dry density, as

specified

in either a standard

(2.5kg

rammer) or heavy (4.5kg rammer) compaction test .

Moisture contení

The aim is to test the specimen under the worst

likely

cond itions that w ill

occur

within

th e

subgrade.

In

practice, soil

is

usu al ly compacted

at

opt im um mois ture content ,

as

specified

in a

com paction test,

an d

then either tested imm ediately

or

soaked

for 4

days

before testing.

Surcharge

weights

Surcharge w eights are placed on the specimen before testing to simúlate the w eight of

pavement m aterials overlying

th e

subgrade.

In

practice,

3

w eights

are usually

used

b ut

this can vary . T he effect of the surcharge weights is more mark ed w ith granular soils .

Testing

top and bottom faces

It is usual A merican practice to test the bo ttom of the specimen w hereas in B ritain both

top and botto m faces are tested and the average take n. Since the top

face

usu ally gives a

lower CB R valué than th e bottom

face,

this variation c an significantly

affect

results.

Method

of

compaction

The AASHTO

specification

stipulates the use of dynam ic compaction

(using

a

rammer

but the BS specification allows the use of

static

compaction (using a

load

frame) or

dynamic compaction (using either

a

ram m er

or a

vibra t ing hamm er) .

Insitu valúes

If

tests

are

carried

out on

completed construction,

the

lack

of

confining

influence

o f the

mould an d drying out of the surface can affect results.

assumption that

CB R

valúes

are

related

to

soil

classification in

some

way.

However , CB R valúes depend not only on soil type b ut also on

density, moisture content and, to some extent,

method

of prepara-

tion. These factors must therefore be taken into account

when

considering correlations

between CBR and

soil cla ssification tests.

A num ber of attemp ts have been made to correlate CB R w ith soil

plasticity. A

correlation b etw een plasticity

Índex

and

C B R ,

for

design

purposes,

is

given

by the

Transport

and

Road Research Laboratory

(1970) ,

as

indicated inTable 7.2. This

is

based

on

wide

e-

perience

of

subgrade soil but is limited to British soils

compacte

at atural

moisture content according to the Ministry of Transport 1969)

specification. Thus, th e precise density and m oisture content condi-

tions

corresponding

to the given CB R valúes is not specified. This

severely limits the use of the

table

outside B ritain.

The valúes used b y the Transport and Road Research Laboratory



C A L I F O R N I A B E A R I N G R A T I O

99

Table

7 2 E S T IM A T E D

L A B O R A T O R Y

C B R

V A L Ú E S

F O R

B R I T I S H S O I L S

C O M P A C T E D A T T H E

N T U R L M O I S T U R E C O N T E N T

C B R

( )

Ty pe

o f

soil

Plasticity

índex

D e pt h o f water table

below

formation level

More than

600mm

600mm o r less

Heavy clay

Silty clay

Sandy clay

Silt

Sand (poorly graded)

Sand

(well

graded)

W ell-graded sandy gravel

70

60

50

40

30

20

10

—

non-plas t ic

non-plast ic

non-plas t ic

2

2

2.5

3

5

6

7

2

20

40

60

1

1.5

2

2

3

4

5

1

10

15

20

ow e

m u c h

to the

w o r k

of

Black (1962),

w ho

obtained correlat ions

between

CBR and

plastici ty Índex

for

various valúes

of liquidi ty

Índex

(defined in

Chapter

6), as show n in

Figure 7.1.

The

valúes obtained

from Figure

7.1

refer

to

saturated soils.

For

unsaturated soils,

the

CB R can be estimated by applyin g a correction to the saturated valué,

using Figure 7.2.

Mor in

and

Todor

(1977)

report

on

attempts

to

correlate soaked

C B R

valúes,

at

op timum mo isture content

and

m á x i m u m

dr y

density

for tropical African and South American soils with the producís:

plasticity

índex

times

th e

perecent passing

the no. 20 0 or no. 40 US

sieves. They concluded that no

well-defíned

relat ionship existed.

However , de Graft -Johnson

et al.

(1969) obtained a correlation of

CBR

with plasticity

and grading

using

th e concept of

suitability

índex,

defined

by:

Suitabil ity

Índex

=

LL.log P/)

where A is the percentage passing a 2.4mm BS sieve. Their fmdings

are given in Figure 7.3. No te, how ever, that the CB R valúes are for

samples compacted

to

máximum

dry

density

at

opt imum mois ture

content according to the

Ghana

standard of compaction. This

specifies

the use of a standard CB R mould and a lOlb

(4.5kg)

ramm er

w i th

an

18-inch (450mm) drop;

to

compact soil

in 5

layers using

25

blows per layer. Samples are tested after a 4-day

soak.




iq ui i ty

índex

in

t í

N;

i» CO CO >

>

o O

O O O O O O* r-* T-

8

7O

6

5O

4

09

O

3

20

1

I 7 i

4

Probable •quilibrium

CBR

under

pavements

in

southern England

1

I I I I

I

1.25

1.3

4 10 40 1OO

California Bearing Ratio

40O

Figure 7.1 Relationships between CBR and plasticity índex at various liquidity

índex

valúes

after Black

1962)

Further work o n lateritic gravéis (de G raft- J o h n s o n

e t al.

1972) led

to the

establishmen t

of a


CBR a n d t he

ratio

of

m á x i m u m

dry

de n s i t y

to

plastici ty Índex

as

s hown

in

Figure 7.4.

Agarwal an d Ghanekar (1970), based on tests of 48 I n d i a n

f ine-grained

soils,

f o un d n o

s ignif icant

correlation between C B R a n d

either

l iquid

l imit ,

plástic l imit

or

plast ici ty Ín dex. How ever, they

d id

obta in better correlat ions w hen opt imum mo isture con tent w as taken

into account .

T he

best

fi t

relat ionship

was for CBR

wi th o p t i m u m

m oi s t ure c on t en t

an d

l iquid l imi t :

T he

soils tested

all

had

C BR

valúes

of less

than

9 and the

s tandard

deviat ion obtained w as 1.8. T hey therefore suggest that th e correla-

t ion is only of sufficient accuracy fo r prel iminary

identif ication

of



CALIFORNIA BEARING RATIO 1 1

100

80

5

60

4

20

o London

Clay

o

Brickearth Harmondsworth

• Black cotton soil Ngong

• Red coffee soil Thika Sagana

Unsaturated CBR = K X

saturated

CBR at

same

moisture

content

0.2

4

1.2

1.4

.6

0.8 1.0

Correction factor K

Figure 7.2 Correction of CBR valúes for

paríial

saíuration after Black, 1962

1.6

120

100

¿ o

8

i

O 6

40

20

O 2 3 4

Suitability índex S

Figure 7.3 Relaíionship beíween suitability Índex a nd soaked CBR valus after de

Graft-Johnson eí

al., 1969}



102


O F SOIL

PROPERTIES

140

2

1OO

O

3

¿C

8O

1 6

o

4

20

i l

1

l l

10

OO 10OO

Máximum

drydensity

kg/m3

Plasticity

índex

Figure

7.4

Relationship between

th e ratio

of

máximum dry densiíy lo plasticity

índex

an d CBRfor laterite-quartz gravéis modified after de Graft-Johnson et al., 1972}

materials. They fur ther suggest that such correlation m ay be of mo re

use if derived for

specifíc

geological regions.

Both th e A A SH T O and

Unifíed

soil

classification

systems were

devised for the specific purpose of assessing the suitability of soils for

use in road and

airfíeld

co nstru ction . Since the C BR valu é of a soil is

also a m easure of its perform ance as a subg rade, logic suggests that

there

should be some general relationship between the soil groups

and C BR v alúes. A p p roxim ate correlations betw een CB R and soil

classes suggested

by íhe US

Highways Research Board

and by the

U S Corps of Engineers are given by Liu 1967) and p resented in

Figures

7.5 and

7.6.

A

similar correlation,

for

South American

red

tropical soils, is given in Figure 7.7.



C A L I F O R N I A B E A R I N G R A T I O 1 3

AASHTO

system

A - 1

A - 1 - a

b

A - 2 - 4

and

5

I

A - 2 - 6 and 7

A-3

A-4

A-5

A -6

and 7

GW

Unified system

em

S P a n

I <

GM

GC

SW

dSM

3P

ML

CL and CH

MH

OL and OH

2

4

6 8

1 15 2 3 4 6 8

Figure 7.5 Approximate relationships between soil classes and CBR valúes

after

Liu, 1967)

GM

[GW

GU

SP

ML CL

I su

sel

M H O L

[CH,OH

3 4

6 8 10 15 20 3O 40 60 80

Figure 7.6 Approximate relationships between

Unified

soil classes and CBR valúes

after

U S

rmy

Corps

of

Engineers, 1970)

A- 2 - 4

A-4

[A-2-6

A-5

A-6

A-7-5

A 7 6

6 8 10 15 20 30 40 60

8 1

150

Figure

7.7 Approximate relationships between

SHTO

soil classes and CBR valúes

for South American red tropical soils

after

Morin and Todor, 1975)



104

C O R R ELA TIO N S

OF

SOIL PROPERTIES

7.3

CBR ND SHE R STRENGTH

The CBR test can be thought of as a bearing capacity problem in

m iniature, in w hich the standard plunger acts as a small

foundat ion.

Terzaghi's bearing capacity equation

for

circular fou ndation s

is:

where

c

Po

B

and

N

is the

cohesión

of the

soil

is

its bulk density

is the overburden pressure at the base of the plunger

is

the

diameter

of the

plunger

N

and

N

are

Terzaghi's bearing capacity factors.

For a saturated

clay

in undrained conditions, the angle of shearing

resistance, < / > (in terms of total stress) is zero. This gives bearing

capacity factors

of

J V

C

=

5.14

2

n),

N

a

=

l

and

N

v

= 0.

Thus,

the

third term in the equation

disappears

and, since overburden pressure

p

0

is

equal only

to the

relatively light pressure exerted

by the

surcharge weights,

the second

term

can

also

be

neglected.

The

equation thus reduces to:

This agrees

with

experience that the number of surcharge w eights

used affects the CBR valué for sands, for w hich

N

q is

much

greater,

but not for

clays.

Using

SI

units,

the CBR

valué

is

100

for a

plunger pressure

of

6900kN/m2 (10001b/in2) at a penetration of 2.5mm, giving:

4

u

x l O O

6900

= 0.09c

where

q

u and c are in kN/m

2

.

Work carried out by Black (1961) on single-sized sand and

correlations with other work fo r clay suggests

that

this approach

gives calculated CBR valúes that are cióse to m easured valúes for field

tests. Lab oratory CBR v alúes can be expected to be higher for sands

because

of the

restraining

iníluence of the

mould. Black (1961) also

sugests that,

when

calculating < j

u

the su stitu on:

c = s tan 0r

is used, where s is the soil suction and < ¿ > r is e true , ngle of internal

friction.

Since,

fo r

cohesive soils

the

true angle

oí

internal friction

can be

estimated from

th e

plasticity índex (see Figure

6.12),

this opens

up the

possibility of predicting both cohesión and CBR valúes from

plasticity Índex and soil suction valúes.



Chapter

SHRINKAGE AND SWELLING

CHARACTERISTICS

Expansiva soils

are those tha t show a marked volume change with

increases and

decreases

of

mo isture

contení.

Such

swelling properties

are restricted to soils containin g clay minerals

which

are susceptible

to penetration of their chemical structure by water molecules.

Clay swelling and consequential g round heave is a common ann ual

phenomenon in áreas where prevailing climatic conditions lead to

signifícant seasonal wetting

and

drying,

th e

greatest

seasonal

heave

occurr ing

in regions w ith semi-arid climates wh ere pronou nced sho rt

wet

and

long

dry

periods lead

to

major moisture changes

in the

soil.

Moisture content changes may also res ult, in these regions and

others,

from

the activities of m an , such as, remov al of vegeta tion and

construction works.

8 1 IDENTIFICATION

The

simplest swelling

identification

test

is

called

th e free-swell

test

(Holtz and Gibbs 1956). The test is performed by slowly pouring

lOcm 3

o f dry

soil <42 5¿m i) into

a lOOcm3

graduated cylinder

fílled

with w ater, and observing the equilibrium swelled volume. Free swell

is

defined

as:

Final

vo lume) — I n i ti al

v o lume)

Free swell = \100( )

Initial volume

Table 8.1 gives

free

swelling data fo r some common clay minerals.

In field situations, the am oun t of swelling or shrinkage , or whe ther

any vo lum e change occurs at

all,

wiíl depend on a num ber o f factors,

such

as

moisture content changes, thickness

of the

deposit, initial

density, groundwater chemistry, confining pressures,

and

possibly

other

factors. However, commonly a fundamental

ingredient

is the

5



10 6

CORRELATIONS

OF SOIL PROPERTIES

Table

8 1

F R E E

S W E L L I N G

D A T A

F O R

C L A Y M I N E R A L S ,

A F T E R M I E L E N Z

A N D

K I N G ,

1955)

Ca-Mont.:

Forest

Mississippi

W i ls on C r ee k D a m ,

Coló

D a vi s D a m , A r iz o n a

,

Osage

W y om in g

(prepared

f rom

N a - M o n t. ),

145

95

45-85

125

N a -M o n t , Osage W y om in g

1 400-1 600

N a-Hectorite , Héctor, California 1 600-2 000

I H ite:

Fithian, I l l inois .

Morris

I l l inois . .

Tazewell, Virginia

Kaolinite:

Mesa

A l ta ,

N e w M é x i co

Macón G e o rg i a

L a n g l e y , N. Carolina . .

Halloysite, Santa R ita , N ew M éxico

115-120

60

15

6

7

Table 8 2 T Y P I C A L R A N G E S O F A T T E R B E R G L I M I T V A L Ú E S

Clay mineral

PL

Dominant pore water catión

Ca2 Na*

LL PL LL

Montmoril lonite

Illite

Kaolinite

65-79

6

26-36

123-177

69-100

34-73

86-97

34-41

26-28

280-700

61-75

29-52

presence

of

m onm orillonite,

or

other smectite,

and

more specifícally

its

propo rtion

in

th e

soil.

In

some instances, clay-mineral type

can

be

identifica from the origin and geological se tting of the soil, tog eth er

with

consideration

of

A tterberg l imits. Typical tanges

of

A t te r b e rg

limits are sh ow n in Table 8.2: note the effect of the dom inant catión in

th e

pore w ater . A nother indicator

of

clay-mineral typ e

is

Skempton's

(1953) activity Ac) which relates plasticity índex

to the

proport ion

of

clay

present in the soil:



S H R I N K A G E A ND S W E L L I N G C H A R A C T E R I S T I C S 107

where

C is the percentage f mer

th an 0.002m m .

Typical activity valúe s

are:

Sodium m ontm or i llon i te 7 .2

Calcium montmoril lonite 1.5

Illite

0.9 and

Kaolinite

0.33 0.46.

8 2 SWELLING POTENTIAL

An indication of the susceptibility of a

soil

to shrinkage or swe lling

due to decreases or increases in m oistu re content is provided by the

swelling potential test.

The

swelling potential

is defmed as the

percentage

swell of a

laterally confined sample which

has

been compacted

to

m á xi m u m

density

at

optimum mois ture conten t according

to the

standard

compaction

test

(BS 1377:1975 Test 12, 2.5kg ram m er m ethod or

AASHTO

T99, 5.51b ram m er m e thod) and then allowed to

swell

under a surcharge of 6.9kN/m2

(llb/in

2

) .

In order to give m eaning to the signifícance of swelling potential

valúes, descriptive

terms

are used for various ranges of swelling

potential, as indicated in Table 8.3.

Tabie J DESCRIPTIVE TERM S

FOR

SWELLING POTENTIAL

Swelling potential ( ) Description

0 1.5

Low

1.5 5 M é di um

5-25 High

25

+

V e r y high

8 2 1 Relation to other properties

The

sw elling poten tial test

is not

no rm ally carried out,

and a

n um b e r

of researchers hav e tried to correlate swelling potential w ith plasticity

índex . Since both the liquid an d plástic lim its and the sw elling

properties of a soil are governed by the am oun ts and types of

clay

minerals present,

it seems

reasonable

to

postúlate that such

a

correlation exists. Seed,

et al

(1962) established th e relationship:

— £r\ ÍDJ\2.4 4



10 8


where

S

is the swelling potential

PI

is the plasticity Índex

and

K

is a

constant,

equal

to 3.6 x

10 ~5 .

This

equatio n applies to soils w ith

clay

contents of between 8 and

65 .

The calculated valué is pro bab ly accura te to

with in

about 33

of

th e laboratory valué. Al th oug h their resul ts are based on w ork wi th

artificial m ixtures

of

sands

and

clays,

the

correlation

has

been show n

to be

applicable

to

na tura l soils. Using

this

equation

and

allowing

for

th e possible 33 error in calculated valúes of swelling potential ,

ranges

of

plasticity

índex

valúes

may be obtained for the

various

classes

o f

sw elling pote ntial ,

as

indicated

in

columns

1 and 2 of

Table

8.4. Also indicated in the table are valúes suggested by Krebs and

Walker

(1971) .

A

correlat ion betw een swel l ing potent ial

and

plast ici ty Índex

w as

found by Chen

(1988),

based on

tests

of 321

undisturbed samples.

He

proposed:

where

=

0.2558

A

= 0.0838

and e is the

natura l num ber , 2 .718.

He

also

established a correlation of plasticity índex againt a

swelling

potent ial obtained

for a surcharge

pressure

of

48kN/m

2

(6.941b/in

2

). A comparison of various correlations between swelling

potent ia l

and

plasticity índex

is

shown

in

Figure 8.1. It should

be

noted that the Holtz and Gibbs

(1956)

correlation given in the figure

is not really comparable with the others s ince their volum e change

measurements

w ere carried out on air-dried specimens of undis turbed

soil.

T he

valúes given

in the chart are

therefore

not

strictly swelling

potential . This is discussed later in this section.

Table 8 4 I D E N T I F IC T I O N

OF

S W E L L I N G S O IL S B S E D

ON

P L S T IC I T Y I N D E X

Swelling potential

Plasticity

índex

Plasticity índex

Low (0-1.5 )

Médium (1.5-5 )

High (5-25 )

Very high (25 +

)

0-15

10-30

20-55

>40

0-15

15-24

25-46

>46

1 Based

on the

relationship given

by Seed el al

(1962).

2 Valúes according to Krebs and W alker (1971).



S H R I N K A G E A N D

S W E L L I N G

CHARACTERI S TI CS 109

1

2

Plasticity índex

-

4

Figure 8.1 A comparison of various correlations between swelling

potential

and

plasticity índex after Chen, 1988)

Although

soils

exhibiting high swelling characteristics usually have

high plasticity Índ ices not all soils with high pla sticity Índices ha ve a

high swelling pote ntial. Thus the plasticity

índex

can be used on ly as

a

rough guide

to

swelling potential.

Logic suggests that th ere should be re la tionships between p otent ia l

for expansión

and

both shrinkage limit

and

linear shrinkage.

Table



110


O F SOIL

P R O P E R T I E S

ble 8.5

S U G G E S T E D C U I D E

TO THE

D E T E R M I N T I O N

OF

P O T E N T I L

FOR

E X P N S I Ó N

US ING S H R I N K A G E LIMIT

A N D

L I N E A R S H R I N K A G E

Potential for expansión

Shrinkage

limit

( )

Linear

shrinkage

( )

Critical

< 1 0

M arg in a l 10-12

Non-critical

>12

>8

5-8

0-5

8 .5

shows

a

general guide

for

these relationships suggested

by

Altmeyer (1955 ) . H owever , a l though

a

knowledge

of

shrinkage limit

is

useful

in assessing

potential

volum e changes, other researchers have

been unable to establish a conclusive correlation between it and

swelling

potential (Chen, 1988).

Work

by

Seed

et al

(1962) suggests that there

is a

correlation

between swelling potential and

trie

contení of clay-sized paríicles

(finer than 0.002mm). Unfortunately, the correlation includes factors

which

depend

on the type of clay

present.

They therefore suggested an

al ternat ive

approach using

the

concept

of

activi ty . Swelling potent ial

is

related

to

activity

as

shown

in

Figure 8 .2 . H owev er , Seed

et al

(1962) suggest that, when using this

figure,

activi ty

be defmed as:

A

Ac~C

This is because a

plot

of plasticity

índex

against clay

content

passes

through the origin for clay contents in excess of 40 but not for lower

clay

contents,

as

indicated

in

Figure 8.3. Using

th e

amended

definition helps to

compénsate

for

this,

for

soils with

the

lower clay

contents.

Holtz and

Gibbs

(1956) correlated volume change with colloid

content (defmed

as finer

than

0.00 I m m ) ,

plasticity Índex

and

shrinkage limit,

as

indicated

in

Figure 8.4. T hey suggest that, because

of

th e uncer ta in ty of the correlations, th e potential for expansión

should

be

assessed

by the

simultaneous consideration

of

all

three

correlations,

as

indicated

in

Table

8.6. Their procedure

has

been

adopted

by the US

W ater

and

Power R esources Service

(formerly the

US

Bureau

of

R eclamat ion) .

It

should

be

remembered that their

volume change measurements, whilst being

m a d e

at a pressure of

6.9kN/m

2 (llb/ in2) are for air-dried undisturbed soils and so are not

directly

comparable

with

th e

valúes

of

swelling

potential discussed

previously (see Figure 8.1). Also, their results are based on only 45

samples.

Figure

8 .5

shows

a

char t given

by

H oltz

and

K ovacs (1981) which



Plasticity Índex

ctivit

'

O-

§

í

a

o

o



112


OF SOIL

PROPERTIES

o

40

32

4

16

8

40

uoiioid

contení

iess

than

O.OOlmm)

- mm

20 40 O 8 16 24

PSasiícíty

¡ndex

Shrinkag»

limit

-

Figure 8.4 Relationships beíween volume change and colloid contení, plasticiíy Índex

and

shrinkage limit, respectively

fo r

air-dry

to

saturated conditions under

a

load

of

6.9kN/m

2

Ipsí)

afíer

Holtz an d Gibbs, 1956)

Table

8 6 E S T I M A T I O N OF P O T E N T I A L V O L U M E C H A N C E S OF C L A Y S

A F T E R

H O L T Z AND

G I B B S 1956)

Data from Índex tests

Colloid

contení

fin r than

O.OOlmm

>28

20-31

13-23

<15

PI

SL

>3 5 <1 1

25-41

7-12

15-28

10-16

<18 >15

Probable expansión

total

volume

change*

>30

20-30

10-30

<10

Potential for

expansión

Very

high

High

Médium

Low

*Based on a loading of

6.9kN/m

2

(llb/in

z

).

gives a guide to the swelling and collapse susceptibility o f soils relate d

to

their liquid limit

and

in-situ

dry

density.

A more sophisticated

relationship

which can take im o account th e

change in moisture content

from

an initial valué to ituí ion is

presented

by

Weston(1980). This correlation,

established

foi soil

in

the Transvaal, is

essentially

a more fully

developed versión

of

previous relationships described by Williams (1957) and Van de

Merwe (1964). Swelling potential

is

given

by:

Swell

( ) =

0.000411

(WLWr4-17 (P)

0.386

1 2 33



S H R I N K A G E

A N D

S W E L L I N G C H A R A C T E R I S T I C S 113

„

18

E

át

16

•o 14

12

1 O O O

8

Expansión

Collapse

2

4 6O

Liquid

H m i t

8

OO

Figure 8.5 A guide to the suscepübility to collapse o r expansión ofsoils,

based

on

liquid

limit

and insitu dry density after

Holíz

and Kovacs,

1981)

whe re

w ¡

is the ini t ia l

moi s tu r e con t en í

P is the

vertical pressure kN /m2) , under which swell

takes

place

a n d W Í S t h e

weighted

liquid limit defmed b y :

<0.425mm\0

where

LL

is the l iquid l imit

8 3 SWELLING PRESSURE

Once

a

potent ia l ly expan sive soil

has

been

identifíed

and a

q ual i ta t ive

indica t ion of the poten t ia l

swell

has been made , an ev aluat ion of the

swelling pressure

is

necessary

for

design purposes. Swelling pressure

can be determined from a one-dimensional oedometer test; a number

of variat ion s of this test ha ve been developed Jennings and Kn ight ,

1957;

Bu rlan d, 1975) but co mm on ly the specimen is

flooded

and the

load required to mainta in

constant

volume is recorded

(Fredlund,

1969).

Alternatively,

th e swelling pressure can be predicted from

empirical relat ionships wi th more rout inely measured parameters .



114

CO RRE L A T I O N S

OF SOIL PROPERTIES

6

• 0 4

•D

C

3

C D

2

0 0

Sweil pressure

<30kPa

Swell pressure

30 125kPa

Sweil pressure

125 300

kPa

Swell

pressure

>300kPa

30

40 6

70

80

Liquid limit

Figure

8.6 Relationship between swell

índex

a nd swelling pressure for a range ofliquid

limit

after Vijayvergiya and Ghassahy,

1973)

ble 8 .7 E S T IM A T I N G P R O B A B L E S W E L L I N G P R E S S U R E A F T E R C H E N , 1988)

Laboratory and

jield data

Percentage

passing

75um siete

>35

60-95

30-60

<30

L

iquid

limit,

( )

>60

40-60

30-40

<30

Standard

penetraí ion

resisíance,

blows¡300mm

>30

20-30

10-20

<10

rnhflhlp

expansión

percent

total

volume

change

>10

3-10

1-5

<1

Swelling

pressure,

kN/m

2

)

>1000

250-1000

150-250

<50

Degree

of

expansión

Very high

High

Méd ium

Low

Vijayvergiya and Ghassahy 1973 ) suggested a means of esí imating

the swelling pressure using a swell índex

(/

s

):

LL



S H R I N K A G E AN D SWELLIN G CH A RA CTERISTICS 115

where w n =

n atural w ater

contení

( ) and

LL = liquid limit.

The

relationship betw een

s an d swelling

p ressure, across

a

r ange

of

liquid limits, is show n in F igure 8.6. Based on experience w ith

expensive

soils

in the

Rocky Mountain

área

of the

United

States,

Chen (1988) suggested a

predictive

relation ship for sw elling pressure

using percentage of fines, liquid limit and the

standard

penetration

resistance,

as

given

in

Table 8.7. No te th at

th e 'probable expans ión

given in Table 8.7 is the sw ell ing poten tial for a coníming load of

48kN/m

2 (10001b/ft2), based

on the

premise that this

is a

typical

foundat ion

pressure

for light

s tructures.

During

th e

past decade

a

number

of

theoretical equations have

been developed for compu ting heave in expan sive soils. Mo st require

an evaluation of the sw elling pressure (Rama Rao an d Fredlun d,

1980; Fredlund et al 1980)

but

some

are

based

on

measurement

of

soil

suction (Snethen, 1980; Johnson, 1980).



Chapter

FROST SUSCEPTIBILITY

Two

potentially damaging

e f fects are

associated wi th frost action

in

soils, the

e xpansión

and lifting of the

ground

in

win te r (frost he aving

and f rost

boiling)

and the

loss

of

be aring capacity d uring

the

spring

thaw.

Soils

that

display

one or

bo th

of these

mani f e s ta t ions

are

referred

to as frost susceptible , The

problem

of f rost

damage

is

wides pread: it

occurs

in te mpérate regions where the re is seasonal soil

freezing as well as in the high lati tude permafrost regions.

9 1

ICE SEGREGATION

Simple

free zing of in ters t i t ial wa t e r causes l i t t le ground uplift .

Frost

heave

occ urs to a

much

greater e xtent where water i s free to enter the

soil

and

migrate

to the

freezing

f ront.

At the

f reezing

front

layers

of

clear ice grow

parallel

to the ground surface by displacing the

overlying

soil layer. The migrating water must come largely

from

groundwater

below

the

layer

in

which

ice is

s egregat ing,

for ice and

frozen

ground

will e í fectively

p r even t

any

downward percola t ion

from

the

ground sur face .

Ic e

segregation

c an

occur ,

not only

where

the

freezing penetrales to saturated soils below the wa te r

table

but

also

when the f reezing front pene trates unsaturate d soils in the

capillary

fringe abo

ve the water table .

The

the rmodynamics

of

mois ture moveme nt

to the f ree zing front

are complex; a useful summary is given by Harris (1987). One

considera t ion

is the prese nce of films of unfroze: adsorbed wa t e r in

frozen

soils, separating soil

ic e f rom

soil

partéele, an d

enabling

particle-free ice lenses to develop (Tagaki

197S

.

Another

is the

concept

of

sec ondary frost he aving whic h involves

the

movement

of

moisture in a frozen fringe abo ve the 0°C isotherm (Miller, 1972;

Konrad

and

Morganstern, 1981). However,

for

practical purposes

the

mec hanism

o f

mois ture movem ent

can be

considered

to be

driven

by suc t ion pressure generated

by ice

g rowth

at the

freezing f ront .

116



FROST

SUSC EPTIBILITY 117

Four

facto rs are of particular

signifícance

in

affecting

the

amount

of

ice

segregation during

soil freezing; the

pore size

of the soil, the

moisture supply,

th e

rate

of

heat extraction

and the confming

pressure. Th eory

and

observation

indícate

that

the

suction poten tial

of

soils and their susceptibility to ice segregation increáses as pore

size

decreases.

However,

th e

low

perm eability o f he avy clays m ay

restrict

water migration sufficiently to prevent

significant

ice segregation

(Penner, 1968).

Thus

highly frost susceptible soils possess pore size

distributions wh ich produce

an

optimum com bination

of

soil su ction

and permeability. In view of the cióse correspondence between pore

size and grain size, and the relative ease w ith which the latter may be

measured,

frost

susceptibility criteria based on soil textures are

frequently used.

9 2 GRA IN

SIZES

The freezing behaviour of soils with varying grain size distributions

has been the subject of m uch s tudy . Beskow (1935) show ed

that frost

heaving increáses rapidly

from

nearly zero

for

coarse sand

to a

máximum in the fine

silt sizes, from which

it

slowly declines

to

approach zero again

in

heavy clays.

For

engineering purposes

Beskow

proposed

a

división

of

soils into

non-frost

susceptible

and

frost

susceptible groups,

and

presented

an empiricaly

derived grading

(Figure

9.1).

This

m ay be simplified to a

general statemen t

that

coarse

and

médium sands

are

generally non-frost susceptible,

that

is ice

lenses do not normally develop w hen they freeze, whereas fine sands,

silts and all but the

heaviest clays

are

frost susceptible

and are

subject

to considerable ice lensing during

freezing,

providing a water supply

is

present. Glossop and Skem pton (1945) observed that

well sorted

soils

in w hich less than 30 of the particles are silt size are non-frost

heaving.

Casagran de (1932) suggested tha t

the

particle size critical

to

soil

heave is 0.02mm : if the proportion of such

particles

is

less than

1 , no

heave

is

expected,

but

considerable heaving

m ay

occur

if

this

amount

is o

ver

3 in

non-uniform soils

and o ver 10 in

very

uniform

soils. The

influence

of the <0.02mm fraction was also

demonstrated

by

Kaplar (1970)

for

gravelly sands w here

the

coarser

fraction was

progressively remo ved. F igure

9.2

shows

th e

relation-

ship between average rate

of

heave

(mm/day) and the

percentage

finer

than 0.02mm; these results were obtained und er

specific

laboratory

conditions

and

they should

only be

used

as a

guide

to the field

response. A qualitative

classification

of frost

susceptibility based

entirely on grain size and used in Sw edish

practice

(Hansbo, 1975) is

given

in Table 9.1.



Average

rate

o

heave

m m d a y

Percent

p

,-«.

o

ti

Ü

C J

«

5

2

2

O

<3

§

sx

3

«>

t

U

S

V;

?

a

a

B

S

» -

» ^

C3-

5

EX

2

u

5

o

í

N



FROST

SUSCEPTIBILITY

11 9

Table 9 1 F R O S T S U S C E P T I B I L IT Y O F

S O I L

G R O U P S : S W E D I S H P R A C T I C E A F T E R H A N S B O ,

1 9 7 5 )

Group

I

II

Frost susceptibility

or

danger

None

Modéra te

Soils

G rave l, sand ,

Fine clay (>4

gravelly tills

0 clay f conten í ) ;

I I I

Sírong

sandy tills, clayey t i lls wi th

>16 fines

1

Silt, coarse clay

(clay

f content

15-25 );

silty tills

f

Defined as 2/j .m.

Defined

sO . O ó m m .

Reed

et al.

(1979) noted that predict ions

from

grain size distribu-

tions failed to take account of the fact that soils c an exist at

different

states

of den sity and therefore porosity, yet they hav e the sam e grain

size

dis t r ibut ion .

They derived expressions for pred icting

frost

heave

Y ,

in m m / d ay ) , and one of their sim pler expressions, based on pore

diameters, is:

Y

=1.694(D

40

/D

80

)-

0.3805

where

D40

and D

80

are the pore diam eters whereby 40 and 80 of

the pores are larger respectively.

9 PLASTICITY

Frost susceptibil i ty tends

to b e a feature of

silty

and

sandy clays, that

is, soils of low to m édiu m plastici ty. Table 9.2 gives a correlation of

Table 9 2 P R E L I M I N A R Y I D E N T I F I C A T I O N O F F R O S T S U S C E P T I B L E S O I L S

Permeability rating

Identification

Frost susceptibility

High permeabi l i ty

Intermedía te

permeabi l i ty

Low

permeabi l i ty

Granular:

< 10 finer

than

5um

G r a n ul a r:

> 10 finer

than

5um

Cohesive:

PI<20

Cohesive:

PI>20

Not susceptible

Susceptible

Not

susceptible



V«ry

Hlfh

Hlfh

Madlum

Low

Very

L ow

30.0

ravolly SAND, SW

Clayey

QRAVEL. QM-QC

QRAVEL,

QM-QC

Loan CLAY, CL

Clay«i

QRAVEL

Sandy

QRAVEL

QP

SANOS

SM-8C

and SC

I l tyQRAVELS

Gravo ly

and

Sandy

CLAYS

CL

SW-SM,

SP-SM

/and

SM

h«av« 1 o O

du«

to

Sandy

QRAVELS

In

« I t u

1920

fraozing

o f

por» water

10

P«rc«ntag fln r than 0-02mm

100

«aturatlon

r

igure 9.3 Average rate ofheê plotted against

per

centage finer than 0.02rv°nfrom

labor atory tests

of a

range of M r

al

soils after Kaplar, 1974



FROST SUSCEPTIBILITY 121

Table 9 F R O S T

S U S C E P T I B I L I T Y

O F S O I L S

R E L A T E D

T O

S O IL C L A S S I F I C A T I O N A F T E R

us

A R M Y

C O R P S O F E N G I N E E R S A N D K R E B S A N D W A L K E R

1971

Group

Description

Fl

Gravelly

soils:

3-20

finer

than

0.02mm

F2

S a n d s : 3-15

finer

than 0.02mm

F3 (a) Gravelly soils:

>20

finer t h a n 0.02mm

(b ) Sands (except silty fine s a n ds ) : > 15 f iner t h a n 0.02mm

(b )

Cla y s:

PI>12

(c) Varved

clays:

w i t h uniform conditions

F4 (a) Silts : including sandy

silts

(b ) Fine silty sands: > 15 finer than 0.02mm

(c)

Lean clays:

PI<12

(d) Varved clays: with non-uniform condit ions

frost

susceptibility and

permeab i l ity w ith grading

and

plasticity

Índex

suitable for preliminary

identification

based on recommendations

by the Tran sp o r t and Road Research Laboratory (1970). A similar

classification sys tem (T able 9.3) involving grading and plasticity w as

established by Linell

et al

(1963) and is used by the U.S. Corps of

Engineers

to

assess

frost

susceptibil i ty

for

pavement design. Once

again th e critical particle size is given as 0.02mm. T he

groups

are in

order of increasing frost suscept ibi l i ty , with group F 4 soils being

particularly

frost

susceptible. A

relationship

show ing the average rate

of heave

(mm/day )

for a

range

of

soil gro up s ,

defined by the Unified

system, is given in Figure 9.3.

Migrat ion of water and frost heaving are

also

influenced by the

mineralogy

of the clay

fraction.

Clay

minerals with

expandable

s t ructures are

able

to

hold

more water but the water is relatively

immobile

compared with

non-expandable

clay minerals. Conse-

q uent l y , strong frost heaving

is

more likely

to be

associated w ith soils

where the fines are devoid of montmori l lomite and related minerals.



R F R N S

AASHTO, 1982. Standard

specifications

fo r transportador materials and methods of

testing

and sampling. American Association of State Highw ay and Trans portatio n

Officials.

ASTM, 1970. Standard test method f or

classification

of

soils

for engineering purposes.

American Association for Testing and M aterials, ASTM Designation D2487-1969.

Abdelhamid, M. S. and Krizek, R. J., 1976. At rest lateral earth pressure of a

consolidating

clay. Proceedings of ASCE Journal ofthe Geotechnical Engineering

División

102:

721-738.

Agarwai,

K. B. and Gh ane kar, K . D., 1970. Prediction o f CBR

from

plasticity

characteristics

of

soils. Proceedings of 2nd South-east Asían Conference on

Soil

Engineering Singapore,

571-576.

Al-Hussaini, M. M., 1977. Contribution to the engineering soil classification of

cohesionless soils. Final repo rt, misce llaneous paper S-77-21, U.S.

A r my

Engineer

Waterways Experimení

Station, Vicksburg, Mississippi, 70pp.

Al-Hussaini, M. M. and Townsend, F. C., 1975. Investigation of K0 testing in

cohesionless

soils.'

Technical Report S-75-11, U.S. Army Engineer Waterways

Experimental Station, Vicksb urg, Mississippi, 70pp.

Altmeyer, W. T.,

1955. Discussion

of

engineering properties

of

expansive

clays.

Proceedings ASCE Journal

of

Soil Mechanics and Foundation División 81 .

Atterberg, A.,

1911.

Lerornas

fórhallande

till

vatten, deras plasticitetsgránser

cch

plastiitetsgrader (The beh aviou r of clays

with

wate r, their limits of plasticity an d

their degrees

of

plasticity). Kungliga Lantbruksakademiens

Handlingar

oc h Tidskrift

50(2): 132-158; also in International Mitteilungen fü r Bodenkunde 1: 10-43.

Azzouz, A. S., Krizek, R. J. and Corotis, R. B., 1976. Regression analysis of soil

compressibility. Soils and Found ations 16(2): 19-29.

BS

1377, 1975. Methods

of

test

fo r

soils

fo r

civil engineering purposes, British

Standards Institution.

BS 5930,1981. Code of practice for site investigations. British Standards Ins titutio n.

BSI

1957. Site Investigation. British Standard

Code of

Practice,

CP

2001.

Beskow, G., 1935. Soil freezing an d frost heaving with special application to roads and

railways. Swedish Geotechnical Society, Series

C, no.

375.

Bishop, A. W., 1958. Test requirem ents of me asuring the coe^cient of earth pressure at

rest.

Proceedings

of Conference on

Earth Pressure

Problt

is, Bru;¿ els, 1:

2-14.

Black, W. P. M., 1961. The calculation of laboratory and i n s i t M valúas of Californ ia

bearing ratio

from bearing

capacity data.

Geotechnique

11 -21

Black, W. P. M., 1962. A method o f estimating the CBR of cohesi „ soils, ai piasticity

data. Geotechnique

12:

271-272.

Bjerrum, L., 1954. Geotechnical properties of Norwegian marim. clays. Geotechnique

4:

49-69.

Bjerrum, L., 1972. Embankments

on

soft ground.

Proceedings

o f

SCE Specialty

Conference

on

Performance

of Earth an d Earth-Supported Structures Purdue

University,

2:

1-54.



REFERENCES 123

Bjerrum, L. and Simons, N.

E.

1960. Comparison of shear stre ngth characteristics of

normally Consolidated

clays.

Proceedings ofthe

ASCE Research Conference on

the

Shear Strength of

Cohesive

Soils

Boulder:

711-726.

Bowles,

J. E ., 1982. Foundation Ana lysis and Design. McG raw-H ill, Singapore, 816pp.

Bozozuk,

M. and

Leonards

G. A .,

1972.

The

Gloucester test

fill. Proceedings ofthe

Earth

and

Earth-Supported Structures Purdue University, 1: 299-317.

Brooker, E.W. and Ireland, H.

O.

1965. Earth pressures at rest related to stress

history. Canad ian Geotechnical Journal 2: 1-15.

Burland,

J. B. 1975. Discussion.

Proceedings

ofóth

Regional

Conference

fo r

África

on

Soil

Mechanics and

Foundation Engineering 1: 168-169.

Campanella, R. G. and Vaid, Y. P., 1972. A simple

K 0- t riaxial

cell. Canadian

Geotechnical Journal 9: 249-260.

Caquot, A. and Kerisel J. 1966. Traite de Mechanique des Sois Gauthier-Villars,

París 509pp.

Cárter, M. 1983. Geotechnical Engineering Handbook. Pentech Press, London, 226pp.

Casagrade, A ., 1932a. Discussion on

frosí

heaving.

Proc.

Highways

Research Board

12 : 169.

Casagrande,

A .,

1932b. Research

on the A tterberg

limits

of

soils.

Public Roads

13(8):

121-136.

Casagrande A., 1948. Classification and identificaíion of soils. Transact ions ASCE

113:

901-932.

Casagrande, A. and Fadum R. E., 1940. Notes

on

soil testingfor engineering purposes.

Harvard University

Gradúate

School of Engineering, pu blication 268.

Castro G. 1969.

Liquefaction

ofSands. Ha rvard Soil Me chanics Series, No. 81 112pp.

Chen F. H. 1988. F oundations on E xpansive Soils. Developm ent in Geotechnical

Engineering

no. 12, Elsevier, Amsterdam, 280pp.

Cou lom b, C.A., 1773 (1976). Essai sur u ne application des regles de max im us et

minimus á

quelques problémes

de

statique,

relatifs á

l architecture. M émoies

de

Mathématique et de Physiqu e, Presentes a l Académie R oyale des Siences par divers

Savns,

et

lüs

dans

ses

Assemblées

París

7:

343-382

(Volume

for

1773 published

in

1776).

de Graft-Johnson, J. W. S. and Bhatia, H. S. 1969. The engineering characteristics of

the lateritic gravéis of Ghana. Proceedings of

7th

International Conference on Soil

Mechanics

and

Foundation Engineering México, 2: 13-43.

de G raft-Johnson, J. W.

S.

Bhatia, H. S. and Yaboa, S. L. 1972.

Influence

of geology

and physical properties on strength characteristics of lateritic gravéis for road

pavemenís. Highway Research Board Record 4 5: 87-104.

ESOPT.

1982.

Proceedings of 2nd European

Symposium

on Penetration Testing

Amsterdam.

Edén W. J.

1971.

Sample

triáis

in overconsolidated sensitive clay. In:

Sampling ofSoil

an d Rock ASTM special technical publication no. 483.

Flaate

K. and

Preber

T.

1974. Stability

of

road embankm ents.

Canadian Geotechnical

Journal

11: 72-88.

Fredlund,

D. G.

1969. Consolidometer test precedural factors affecting

swell

proper-

ties.

Proceedings

of 2nd

International

Conference

on

Expansive Clay Soils Texas

435-456.

Fredlund

D. G. Hasan J. U. and Filson H. L.

1980.

The

prediction

of total

heave.

Proceedings

of

4íh

International Conference on Expansive Soils 1-17.

Gibbs H.J., 1969. Discussion on expansive soils and m oisture movement in partially

saturated soils. Proceedings of7th International

conference

on Soil Mechanics and

Foundation

Engineering.

México City.

Gibbs H. J. and Holtz W.

G.

1957. Research on determining the density of sand by



124 C O R R E L A T I O N S

OF SOIL

PROPERTI ES

spoon

penetration

testing.

Proceedings of

4th

International

Conference on Soil

Mechantes and

Foundation Engineering

London,

1: 35-39.

Gibson,

R . E.,

1953. Experim ental

determ ination of the

tru e cohesión

and

true angle

of

internal

friction

in clays. Proceedings of 3rd International Conference on Soil

Mechanics

and

Foundation Engineering Zurich, 126-130.

Glossop,

R. and Skem pton, A. W.

1945. Particle size

in

silts

an d

sands.

J.

Inst.

C ivil

Engineers 25:

81-105.

Gregg,

L. E., 1960. Earthworks. In: K. B. Woods Editor), Highway Engineering

Handbook, McGraw-Hil l .

Hansbo,

S.,

1975.

Jordmatariallára.

Almquist

an d

Wiksell Fórlag

AB,

Stockholm,

218pp.

Harrison,

J. A.

1988. Using

the BS

cone penetrometer

for the

determination

of the

plástic limit of soils. Geotechnique 38 :

433—438.

Harris, C. 1987. Mechan isms of mass mov em ent in periglacial env ironm ents. In: Slope

Stability

Eds. M. G. Anderson and K. S. R ichards) , John Wiley, 531-560.

Hazen, A., 1911. Discussion of Dams on Sand Foundations. Transactions ASCE 73:

199-203.

Hilf,

J.

W.,

1975. Compacted

Fill. In: H. F.

Winterkorn

and H. Y.

Fang

Editors) ,

Foundation Engineering Handbook. V an N os t rand Reinholcí.

Holtz, R. D. and Broms, B. B., Sweden.

Proceding

o f ASCE Speciality Conference

on

Performance of

Earth

and Earth-Supported Structures Purdue Universi ty, 1:

435-464.

Holtz, R. D. and Holm, G., 1979. Test em ban km ent on an organic silty clay.

Proceeding

oflth

European

Conference on

Soil Mechanics

and

Foundation Engineer-

ing

Brigh ton , 3: 79-86.

Holtz, R. D. and Kovacs, W. D., 1981. An Introduction to

Geotechnical

Engineering.

Prentice-Hall,

New Jersey, 733pp.

Holtz, W. G. and Gibbs, H .

J.,

1956. Engineering properties of expansive clays.

Transactions

ASCE 121:

641-677.

INSITU.

1986. Proceedings

o f

ASCE Speciality

Conference.

INSITU

86 ,

Blacksburg.

ISOPT. 1988.

Proceedings of Is t International Symposium on Penetration Testing

Orlando.

Jennings, J. E.B. and Knight, K. 1957. The

prediction

of total heave

from

the double

oedometer test. Trans.

S.

Afr. Instn. Civ. Engrs.

7:

285-291.

Johnson, L.

D.,

1980. Field test sections on expansive

soil.

Proceedings of 4th

International Conference on Expansive Soils.

Joslin, J.

G.,

1959. Óhio s typical mois ture - de nsity curves. Symposium

on

A pplication

of

Soil Testing in Highway Design an d Construction

ASTM Special

Technical

Publication, 239: 119-123.

Kaplar,

C .

W., 1974. U.S.

Army

Cold Regions Research Laboratory, CRREL Report ,

250.

Karlsson,

R.,

1977. Consistency Limits. Swedish Council for

Building

Research,

document D6,

40pp.

Kenney, T. C., 1959. Discussion of

Geotechnical properties

of glacial lake clays.


of

Soil Mechanics

an d

Foundation División 85:

67-79.

Kenney,

T.

C., 1976. Formal

and

geotechnical characteristics

of glacial-lake carved

soils.

Laurits

Bjerrum Memorial

Volume -

Contributions

to

Soil Mechanics,

Norwegian

Geotechnical Institute, Oslo,

15-39.

Kezdi, A ., 1974. Handbook of Soil Mechanics

Vol.

J: Soil Physics. Elsevier.

Konrad, J-M.

and Morgenstern, N. R.,

1981.

The

segregation potential

of a

freezing

soil . Canadian Geotechnical Journal,

18:

482-491.

Krebs,

R. D. and

Walker,

E.

D., 1971.

Highway Materials.

McGraw-Hil l .



REFERENCES 125

Publication

272, Department

of

Civil Engineering, Massachusetts Institute

of

Technology, 107pp.

Ladd C.

C .,

1975. Foundation Design

of Embankments

Constructed on Connecíicut

Valley

Varved Clays. Research Report R75-7, Geotechnical Publication 343,

Department of Civil Engineering, Massachusetts Institute of Technology, 438pp.

Ladd

C. C. and

Foote,

R.,

1974.

A new

design procedure

fo r

stability

of soft

clay.

Proceedings

ASCE Journal ofthe Geotechnical Engine ering División, 100: 763-786.

Ladd,

C.

C. Foote

R.,

Ishihara,

K.

Schlosser,

F. and

Poulos,

H. G.

1977.

Stress-deformation

and

strength characteristics. Proceedings of 9th International

Conference on Soil Mechan ics and

Foundation

Engineering, Tokyo 2: 421-494.

Ladd,

C. C. and

Luscher,

U.

1965. Engineering p roperíies

o f t he soils underlying the

M.I.T. campus. Research repo rt R65-68, Dep artmen t of Civil Engineering, Mass-

achusetts Institute

of Technology.

Ladd, R.

S.

1965. Use of

electrical

pressure transducers to measure

soil

pressure.

Research Report R65-48. Soil Publication 180, Department of Civil Engineering,

Massachusetts Institute

of

Technology, 79pp.

Lambe, T. W. and Whitrnan, R. V., 1979. Soil Mechanics. John Wiley, New York ,

553pp.

Lañe

K. S . and Washburn, D.

E.

1946. Capillary tests by capillorimeter an d

soil

filled

tubes. Proceedings Highway Research Board.

La

Rochelle,

P.

Trak,

B. Tavenas F. and

Roy ,

M.

1974. Failure of

a

test embank ment

on

a sensitive

Champlain

clay deposit.

Canadian

Geotechnical

Journal, 1:142-164.

Leonards

G. A ., 1976. Estimating consolidation settlements of

shallow

fo undations on

overconsolidated

clays,

Tran spo rtatio n Research Board , special report 163.

Leonards,

G . A. and

Girault,

P.

1961.

A

s tudy

o f t h e

one-dimensional consolidation

test. Proceedings

of the 5th

International Conference

on

Soil Mechanics

an d

Foundation Engineering,

París

1: 116-130.

Liao,

S. C. and

W hi tman,

R. V.

1986. Ov erbu rden Correction Factors

for SPT in

Sand.

Proceedings

ASCE Journal of Geotechnical

Engineering

División,

112:

373-380.

Liu,

T.

K.

1967.

A

review

o f

engineering

soil classification systems. Highway Research

Board Record, 156: 1-22.

Lowe,

J.

Zacchea

P. F. and

Feldman,

H.

S. —964. Consolidation testing wi th back

pressure.

Proceeding

ofASCE Journal ofSoil

Mechanics a nd

Found ation División, 90 :

69-86.

Massarsch K.

R.,

1979. Lateral earth pressure in normally consolidated clay.

Proceedings

of7th

European Conference on Soil Mechan ics and Found ation Engineer-

ing,

Brighton, 2: 245-250.

Menard, L. and Rousseau, J. 1962. L evalu aíion des tassements tendances nouv elles.

Sols-Soils,

1: 13-28.

Menzenbach,

E.

1967. Le capacidad suportante de pilotes y grupos de pilotes.

Technologia Ingeneria Civil), Series

2, No. 1.

Mesri

G.

1973. The

coefficient

of secondary compression.

Proceedings

ASCE Journal

ofSoil Mechanics an d Foundation División, 99 : 123-137.

Mesri,

G. and

Godlewski,

P.

M. 1977. Time

and

stress compressibility interrelation-

ships. Proceedings ASCE Journal o f Geotechnical Engin eering División, 103:

417- 30.

Meyerhof, G. G. 1936. Penetration tests and bearing capacity of cohesionless soils.


o f

Soil Mechanics

an d

F oundation División, 82.

Meyerhof, G. G. 1974. General Report: outside Europe. Proceedings of

European

Symposium on Penetration Testing,

Stockholm

2: 40-48.

Mielenz R. C. and King M.

E. 1955. Physical-chemical properíies

and

engineering

performance of clays. Calif. Div. Mine s Bull., 196-254.

Miller

R. D.

1972. Freezing

and

heaving

of saturated and

unsaturated

soils. Highways

Research

Record, 393: 1-11.




Mil l igan

V . 1972.

Discussion

of E m b a n k m e n t s on

soft

g r o u n d . Proceedings

ofASCE

Specialty Conference on Performance of Earth and Earth-Supported Structures.

Purdue Universi ty

3: 41-48.

Ministry of

Transport

1969. Specificaíion fo r road and b r idge works . HMSO.

Mitchel l J. K. and Gardner W. S. 1975. In-si tu measurement of

v o l u m e

change

characteristics. Proceedings ofthe ASCE Specialty

Conference

on

In-situ measure-

ment ofSoil Properties. Raleigh N. Carol ina 2: 333.

Mitchell J. K. 1976.

Fundamental

of

Soil Behaviour.

John Wiley 422pp.

Mor in W. J. and

Todor

P.C. 1977. La ten tes and laíeritic soils and other problem

soils

o f t h e

t ropics. U ni ted States Agency for Interna tional Developmen t AlO/csd

3682.

Peck R. B. and B azaraa A. R. S .

S.

1969. D iscussion. Proceedings ASCE Journal of

Soil Mechanics and Foundation División, 95:

305-309.

Peck R. B. H a n s o n W. E. and Thorn burn T . H . 1974. Founda tion Engineering. John

Wiley London

514pp.

Pen ner E. 1968. Pa rticle size as a basis for pred ictin g

frost

action in soils. Soils and

Foundations 8: 21-29.

Poulos

H. G. and

Davis

E.

H.

1974. Elastic Solutions for Soil and Rock Mechanics.

John Wüey N ew Y o r k 41ípp.

Ra ma Rao R. and Fredlund D.G. 1988.

Closed-form

heave

solutions

for expansive

soils. Proceedings

SCE

Journal ofGeotechnical Engine ering División, 114: 573-588.

Rankine W. J . M. 1857. O n

the

stab il i ty of loóse earíh.

Philosophical Transactions of

the

Royal

Society,

Lon don 147.

Reed M . A .

Lovel

C . W . Alíschaeffi A. G. and Wood L. E. 1979.

Frost

heav ing r a te

predicted from pore size distribution.

Canadian Geotechnical

Journal, 16: 463-472.

Robertson P. K. and Companel la R. G. 1983. Interpre tat ion of cone pene trat ion tests

Part

I:

Sand.

Canadian Geotechnical Journal, 20: 718-733.

Roscoe K. H.

Schofield

A. N. and Wroth C. P. 1958. On the y ielding of soils.

Geotechnique,

8:

2-52.

Seed

H .

B.

W oodw ard R. J . and Lu n dg r a n

R. 1962. Prediction

of

swelling po tential

of

compacted clays. Proceedings ASCE Journal of Soil Mechanics and Foundation

División,

88: 107-131.

Seed H. B. 1976. Evaluat ion of soil l iquefaction effects on

level

g round during

ear thquakes . SCE Speciality Session, Liquefaction Problems

in

Geotechnical

Engineering, Prepr int 2752 ASCE Nation al Conv ention

Sept./Oct.

1976

1-105.

Simons

N. E. 1958. Discussion of test requirements fo r m easuring th e coefñcient of

earth pressure at rest.

Proceedings

of Conference on

Earth pressure Problems,

Brussels

3:

40-53.

Singh R. Henkel D. J. and Sangrey D. A. 1973. Shear an d K

0

swelling of

overconsolidated clay. Proceedings ofSth International

Conference on

Soil Mechan-

ics

and Foundation Engineering, 367-376.

Skempton

A .

W.

1944. Notes

on the

compressibil i ty

o f

clays. Qu at. Journ. Geol. Soc.,

100:119-135.

Skempton

A .

W. 1953.

Th e

colloidal activity

of

clays.

Proceedings

of 3rd

International

Conference

on Soil Mechanics and Foundation •qineering, Zur ich 57-61.

Skempton A. W. 1954. The pore pressure

coeL.

. snts A and B.

Geotechnique

4:

143-147.

Skempton A. W. and Nor they R. D. 1952. The sensitivity o f clays. G eotechnique, 3 :

30-53.

Skempton A. W. and Bjerrum L. 1957. A contribution to the settlement analysis of

founda t ions

on clay. Geotechnique, 7 : 168-178.

Ske mp ton A. W. 1986. Standa rd penetration test procedures and effects in sand of

overbu rden pressure relative density particle size ageing and overconsolidation.

Geotechnique

36:

425-447.



REFERENCES 127

Snethen D. R. 1980. Characterization of expansiva soils using soil suction data.

Proceedings

of 4th

International

Conference on

Expansive Soils 1:

54-15.

Sowers

G.

F.

1979.

Introductory Soil Mechanics

and

F oundations.

Macmillan.

Tagaki

S.

1979. Segregation

freezing as the

cause

of

suction forcé

for ice

lens

format ion.

Engineering Geology 13 :

92-100.

Tavenas

F.

Lebland

P.

Jean

P.

and

Leroueil

S.

1983a.

The

p ermeability

of

na ture

soft clays.

Part

I:

Methods

of

laboratory measurement. Canadian Geotechnical

Journal

20: 629-644.

Tavenas

F.

Jean P. Lebland P. and Leroueil S. 1983b. The permeability o f na tura l

soft clays. Part II: Permeability characteristics.

C anadian Geotechnical Journal

20:

645-660.

Tavenas

F. and

Leroueil

S.

1987. State

of the art on

laboratory

an d in-situ

stress-strain-time behaviour of clays. Proceedings of International

Symposium

on

Geotechnical Engineering

of Soft

Soils México City

1-46.

Taylor

D. W.

1948.

Fundam entáis

ofSoil

Mechanics.

John Wiley

N ew

Yo rk 700pp.

Terzaghi K. and Peck R. B. 1967. Soil Mechanics in Engineering Practice. John

Wiley

London 729pp.

Teng W. C. 1962. Fou ndatio n Design. Prentice Hall Englewood

Cliffs

N.J.

Tho rbu rn S. 1963. Tentative correlation chart for the standard penetration test in

non-cohesive soils.

Civil Engineering Public

Works

Review.

Tok imats u K. and Yo shim i Y. 1983. Em pirical correlations of soil liquefactio n based

on

SPT N-values and fines con ten í .

Soils

and

F oundations

23:

56-74.

Tomlinson M. J. 1980. Foundation Design and Construction. Pitman Lon don

793pp.

Transport and Road Research La bora tory 1970. A guide to the structural design of

new

pavements. TRRL Road Note

29 HMSO.

U.S. Army Corps of Engineers 1970. En gineering and design: pavement design for

frost

conditions. Corps of Engineers EM-110-345-306.

U.S. A rm y Engineer Waterways Exp erimen t Station 1960. The

Unified

Soil Classifica-

tion System. Technical memorándum

No.

3-357.

U.S. Bureau

of

Reclam ation 1974. Earth Manual. Denver

810pp.

U.S. Federal Av iation Agency 1984. Ai rpo rt Paving. Advisory Circular

150/5320-6.

U.S. Navy 1982. Design Manual: Soil Mechanics Foundations and Earth

Structures

N avy

Facilities Engineering Command

Navfac

U.S. Naval Publications and

Forms Center.

V an der Merwe D. H. 1964. Prediction of heave from the plasticity Índex and

percentage

clay

fraction

of

soils. Civil Engineer in South

África

6:

103-107.

Vijayvergiya

V. N. and Ghazzaley O. L 1973. Prediction of swelling p otential fo r

natural clays. Proceedings

ofSrd

International

Conference on

Expansive Soils

Haifi

1:227-236.

Wallace G. B . and Otto W . C. 1964. Differential settlement at Selfridge A ir Forcé

Base.

Proceedings ofASCE Journal

ofSoil

Mechanics

and

Foundation División

9 :

197-20.

Weston D. J. 1980. Expansive roadbed treatm ent for Southern África. Proceedings of

4íh International Conference on

Expansive Soils.

1:

339-360.

Williams

A. A. B.

1957.

Discussion.

Trans. S. Afr. Instn. Civ. Engrs.

8.

Wo ods K. B. and Litehiser R .R . 1938. Soil mech anics applied to highway engineering

in

Ohio.

Ohio

State Univ ersity Engineering Experimental Station Bulletin 99.

Wroth

C . P.

1972. General theories

of

earth pressures

and

deformation. Proceedings

of5th European

Conference

on

Soil M echanics

a nd

Foundation Eng ineering Madrid

2: 33-52.

Wroth C. P. and Wood D. M. 1978. The correlation of índex properties with some

basic engineering properties of soils. Canadian G eotechnical Journal 15:

137-145.

Wroth

C.

P.

1984.

The

interpreta tion

of

insitu soil tests.

Geotechnique

34:

449-489.



IN X

AASHTO soi l

classification

system 14

21 27 34 35

and CBR valúes 102

compared

w i th th e Uni f íed

system

37

38

AASHTO standard compact ion tests

44

Activity 11

and expansive m inerals 107

and plasíiciíy

índex

10 6

and swell ing potent ial 110

Adsorpt ion

complex

4

Angle of in ternal friction 12 89

Angle

of

shearing resistance

12 76 89

ASTM/Unif ied soil classif icat ion

system

14

and CBR valúes 102

and

frost susceptibiliíy

121

se e

also Uni f íed

soil classification

system

Atterberg

limits

see Consistency limits

BS soil classification system 14 17

27-29

BS soil descriptions 17

BS s tandard compaction tests 44

Bulk

density 39

California Bearing ratio

2 97 98

and liquidity Índex 99

and m áx im um

dry

density 99 100

and

opt imum moisture content

100

and

plast ici ty Índex

98 100

and

shear strength

104

an d soil classification 102

and suitability índex 99

Casagrande soil classification

system

14

Cations

223

Classifícation systems fo r soils

review 13 14

for

frost susceptibility

119 121

see a lso under AA SHT O BS

Unif íed

systems

Collapse potent ial

and densi ty

111

112

Coefficient

of

c omp ressibil i ty

56 57

typical

va lúes

61

Coefficient

of cu rva t u re 17

Coeffícient

of earth pressure

92-96

active

92 93 95

passive 92 93 95

at rest 95

Coeffícient of

p ermeabi l i ty 5 0 5 1

and consolidat ion 65

and

grading 51 53

and soil classification 51

typical

valúes

51

Coefficients of secondary

con solidation 68 69

Coefficient

of

uniformity

17

Coefficient

of

volume

compressibility

56 57

Cohesión 6 76-78

Cohesión soils 4

Compacted densi ty 43^47

and CBR 99 100

and shear st rength 81

Compaction tests 43^45 49

Compressibility 55

Coefficient of 56 57 61

coefficiení of

vo lum e

56 57

Compression Índex

58

modified 58

valúes and corre la t icns 60

Consistency limits . 6 7

and consolidation

11

an d expansiveness 106

and shear st rength 11

se e

also Liquid

Plástic a nd Shrinkage

limits

Consolidation 2 55

and consistency limits 10

and compressibility 65

128



and pe rmeabi i i t y 65

coeff ic ien í

o f

65-68

pa r a m e í e r s 55-58

t heory 58

Conso l ídome íe r

55

C o n s t r a i n e d m o d u l u s 60

Drained

shear

s t r en g t h

see shear s trength

D e f o m a t i o n m o d u l u s

60

Dry

density

39

Effec t ive

shear

s t reng íh

se e shear

s í r eng th

Effec t ive

stresses

76, 78-80

E x p a n s i v e

soils 11,

12,

105-107

Free

swell 105

Frost

heave

11 9

Frost susceptibility 116 11 7

and g rad ing 117-119

and plasticiíy

Índex

119-121

and soil classificaíion

119,

121

iden í i f í ca t i on o f

soil

119

Grad ing

1-3

and frost suscept ib i l i ty 117-119

and pe rmeabi i i ty 53

classifications 4

effects o n

o the r prope r í i e s

2

Hazen s fo rm ula

53

Hveem s íabi lometer

116

97

Ice

segregaíion

Ilute

107

In te rna l f r i c t i on , angle

o f

Kaol in i t e 107

1 2 , 8 9

Lateral earth pressures

92-96

se e also coefficient o f

earíh pressure

Linea r shr inkage

and

swelling

poíential 110

Liqu id i ty Índex

82

and CBR 99

and

shear stren gth 81-84

and

sens i t ivi ty

83

Liquid

l imit 6-8 10-12

and CBR 99

and swe l ling po ten t ia l 1 10

and swell ing pressure

115

M á x i m u m d r y

dens i íy

45

and CBR

99,100

a n d o p í i m u m

m o i s í u r e c o n t e n t

46

and s he a r s t r eng th 81

s t a n d a r d cu rves f o r 49

M o d i f i e d compress ion Índex

58

M o i s íu r e c o n t e n t

and

swell ing p o t e n t i a l

112, 11 3

Mois íure-densi ty curves , íypical 49

M o n t m o r i l l o n i t e

106, 107,

121

O e d o m e í e r

55

O p t i m u m m o i s t u r e c o n t e n í

45

and CBR 100

an d

m á x i m u m

d ry

d ens i ty

46

and

p las t i c i íy

46

íypical

mo is lu re -dens i íy cu rves

49

Overconso l ida íed

c lays

86, 87

Parl ic le s ize d is t r ibut ion

see

Grading

Perm eab i l i í y 2

and

conso l ida l ion

65

and grading 51, 53

and

soil classification

51

coeff ic ient of 50, 51

Plast ic i íy 3, 6

Plasíiciíy

í ndex

7, 11

and

a c í iv i íy

106

and CBR 98, 100

and

f rosl susceplibi l i íy

119-121

an d swelling poíeníial 107, 112, 113

Plástic limit 6-8 10-12

and op t imum moisture co n tent 46

Píate

bear ing

tesí

74, 75

Poisson s raíio 60, 73

Relaí ive

dens i íy

40

Secondary com press ion

55

coefficienís of 68, 69

Sensiíiviíy 83

Seítlement 58, 59

co r rec í ions 6 2 , 6 5

of

sands

and

gravéis

70-75

Shear ing

resisíance, angle

of 76

Shear s í rengíh 2

76-92

and CBR 104

and cons is íency l imiís 10

and

l iqu id i íy

í ndex 82

and SPT va lúes 88



P R C

i K í ic,b

and s we l í ing pote ní ia l 110

dra ined 89

eífective

89, 90

of ciays

89, 90

of granular soi ls 90-92

parameters 76

r e m o u l d e d

81-83

tota l and

e ffecí ive 78-80

u n d r a i n e d

80-88

Shrinkage l imit 6 , 9-11

Sieve

analys is

se e

Grading

Sieve

sizes

3

Smectite 106

Soil c lass i f icat ion sys tems,

see

Class i f icat ion sys tems

see

also under AASHTO,

BS,

U n i f i e d

systems

Stabil i ty

analysis

79, 80

Standard compact ion te s i s 43-45

one

p o i n t

test 49

Standard peneíra í ion

test

40-43,

70-72

and seí t leme ní 71,

72

and

undrained shear

s t r eng t h

8 8

Suct ion pressure

116

S ui íab i l i ty índe x 99

Swell Í n d e x 11 4

Swel í ing p o t e n t i a í Í 0 7

and d e n s i í y 111, 112

and l inear shrinka ge 110

and l iqu id l imit 112

and

m oi s t ure c o n t e n í

112-,

113

and

p la s t i c i íy Índe x 107-109,

112

and

shr inkage l imi t

110,

112

and vert ica l pressure

112,

113

Swel í ing

pressure 113-115

and l iqu id l imit 115

an d SPT

va lu é

115

and swel l

Í nde x

114

Tota l and

efiective stress

analysis 78-80

76

Undra ined shear s t reng íh

see

shear s í rengíh

*Unified soi l c lass i f icat ion sys tem

and CBR

v a l ú e s

102

and frost

suscept ibí l i ty

121

compared w i t h o íher sys tems

Y o u n g s m o d u l u s

59

4

8

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