ORIGINAL PAPER

Slope Stability Problems and Back Analysis in Heavily JointedRock Mass: A Case Study from Manisa, Turkey

Mutluhan Akin

Received: 13 February 2012 / Accepted: 1 May 2012 / Published online: 24 May 2012

� Springer-Verlag 2012

Abstract This paper presents a case study regarding

slope stability problems and the remedial slope stabiliza-

tion work executed during the construction of two rein-

forced concrete water storage tanks on a steep hill in

Manisa, Turkey. Water storage tanks of different capacities

were planned to be constructed, one under the other, on

closely jointed and deformed shale and sandstone units.

The tank on the upper elevation was constructed first and

an approximately 20-m cut slope with two benches was

excavated in front of this upper tank before the construc-

tion of the lower tank. The cut slope failed after a week and

the failure threatened the stability of the upper water tank.

In addition to re-sloping, a 15.6-m deep contiguous

retaining pile wall without anchoring was built to support

both the cut slope and the upper tank. Despite the con-

struction of a retaining pile wall, a maximum of 10 mm of

displacement was observed by inclinometer measurements

due to the re-failure of the slope on the existing slip sur-

face. Permanent stability was achieved after the placement

of a granular fill buttress on the slope. Back analysis based

on the non-linear (Hoek–Brown) failure criterion indicated

that the geological strength index (GSI) value of the slope-

forming material is around 21 and is compatible with the in

situ-determined GSI value (24). The calculated normal–

shear stress plots are also consistent with the Hoek–Brown

failure envelope of the rock mass, indicating that the

location of the sliding surface, GSI value estimated by back

analysis, and the rock mass parameters are well defined.

The long-term stability analysis illustrates a safe slope

design after the placement of a permanent toe buttress.

Keywords Slope stability � Back analysis � GSI �Non-linear failure criterion � Water storage tank �Heavily jointed rock mass � Retaining pile wall �Block punch index test

1 Introduction

The potable water supply of a settlement is usually stored in

water tanks of different capacities. The dimensions of a tank

are related to the water demand, calculated with respect to

the population. A water tank both regulates the water

pressure in the network and reserves a water supply trans-

mitted from the source location. Furthermore, a water tank

should be adequately elevated in order to fully maintain the

hydraulic pressures required for potable water network

distribution. Thus, water storage tanks are mostly located on

hills or uneven terrain, and a cut slope is usually excavated

so as to construct the concrete tank on a flat surface.

Although slope stability problems concerning water storage

tanks are not very common during the construction or post-

construction period, fatal events may occur after such

incidents (Calderon et al. 2009). Water leakage from the

tank may considerably reduce the shear strength of the slope

material, leading to slope failures and catastrophic acci-

dents. Moreover, excavations to create cut slopes during

construction may trigger slope instabilities, which may also

affect the safety of nearby structures. The failure of slopes

and the substantial costs of remedial measures are mostly a

consequence of unsatisfactory geological and geotechnical

investigations and inadequate interpretation of acquired

data during preliminary design (Lee and Hencher 2009).

The construction of two reinforced concrete (RC) water

storage tanks started at the end of 2005 in Manisa, Turkey,

to store a portion of the potable water demand. The

M. Akin (&)

Department of Mining Engineering, Yuzuncu Yil University,

Zeve Campus, 65080 Van, Turkey

e-mail: mutluhanakin@gmail.com

123

Rock Mech Rock Eng (2013) 46:359–371

DOI 10.1007/s00603-012-0262-x

location map of the study area is shown in Fig. 1. The

latitude and longitude of the construction site are

533338.27 E and 4273272.91 N. The capacities of the tanks

are 3,000 and 7,500 m3, respectively. Both tanks were

constructed, one under the other, on steep terrain (slope

inclination C30�). The horizontal distance between the two

tanks is approximately 30 m. In addition, the elevation of

the 3,000-m3 tank (WT1) is 153.3 m, whereas the 7,500-m3

tank (WT2) is situated at 133.8 m. A general cross-section

of the construction area with the location of the two RC

water storage tanks is provided in Fig. 2. It should be noted

that the construction area was restricted by an expropria-

tion boundary and an unusually steep cut slope had to be

excavated to place the two structures in a tight area, due to

space limitations. After the excavation, the cut slope failed

and the failure affected the stability of WT1. Several

effective and ineffective remedial works were carried out

to retain the failed slope and WT1.

In this paper, the repeated failure of a cut slope and a series

of remedial works for slope stabilization are explained. Fur-

thermore, the slope failure which occurred after the con-

struction of the upper tank is back-analyzed to assess the shear

strength parameters of the slope-forming material. In addition,

the movements monitored after the construction of a retaining

pile wall is evaluated and, finally, the efficiency of the

placement of a granular fill buttress on the slope is analyzed.

2 Geology of the Study Area

The study area is situated on highly fractured and deformed

rocks of the Bornova Flysch Zone. The flysch zone com-

prises large blocks of Mesozoic limestone, basalt, serpen-

tine, and radiolarian chert with a highly disturbed clastic

matrix of Cretaceous to Paleocene age (Okay and Altiner

2007). Moreover, Neogene-aged yellowish brown marl

layers crop out in the same zone. The foundations of the two

water storage tanks as well as the cut slope exist in the

clastic matrix of the Bornova Flysch Zone. It consists of

gray graphitic shale and alternating beds of sandstone and

shale units (Fig. 3a, b). These units are closely jointed,

sheared, and folded with a chaotic structure (Fig. 3c). There

is no observed groundwater table in the study area, except

local wet zones after heavy rains. As the above-mentioned

geological units exhibit similar geotechnical properties, it is

quite difficult to differentiate the exact unit boundaries in

the construction area. The discontinuity spacing of the slope

material is mostly between 5 and 10 cm. As a particular

note, the discontinuity surfaces are usually smooth and

soapy, which drastically decreases their shear strength,

especially during heavy rains. There is no certain discon-

tinuity orientation, as the rock mass is heavily jointed. In

this highly deformed material, a circular slope failure is

more probable than a structurally controlled instability,

since there is no distinct discontinuity surface on which

failure can occur (Anderson and Richards 1987; Ozdemir

and Delikanli 2009; Sharifzadeh et al. 2010). The sliding

surface in heavily jointed rock masses involves both natural

discontinuities aligned on the sliding surface and some

shear failure through intact rock (Wyllie and Mah 2004).Fig. 1 Location map of the study area

Fig. 2 A general cross-section of the construction area (note: the

vertical scale is exaggerated)

360 M. Akin

123

3 History of the Slope Instabilities

A detailed geotechnical survey was not carried out on the

construction site during preliminary investigations in this

project. The physico-mechanical properties of the litho-

logical units were estimated by observational studies and

literature data without any laboratory or in situ tests.

In the preliminary stage of the project, the construction

of the 3,000-m3 tank (WT1) at the higher elevation was

started. When the tank construction was about to be com-

pleted, an approximately 20-m high cut slope with two

benches was excavated in front of WT1 to make room for

the foundation of the 7,500-m3 water storage tank (WT2) at

the lower elevation (Fig. 4). Having completed the slope

excavation, the foundation of WT2 and a drainage ditch

along the tank perimeter were excavated. One week after

excavation of the drainage ditch, the cut slope in front of

WT1 instantly failed. Major and progressive tension cracks

at the top bench and a small-scale horizontal movement at

the toe were observed after the failure (Fig. 4). The slip

surface shown in Fig. 4 is estimated considering the main

tension crack and the horizontal movement at the slope toe.

Eventually, the slope failure threatened the stability of

WT1. Also, a separation of several millimeters in scale

between the main tank and the maneuver room sections of

WT1 occurred and was monitored (Fig. 5).

As an immediate remedial measure to prevent slope

failure and to protect the stability of WT1, a granular toe

buttress was constructed and re-sloping was performed by

removing slope material from the crown to lower the

sliding forces (Fig. 6). Further slope movement was pre-

vented by the above-mentioned temporary remedial mea-

sures. However, the toe buttress covered a large portion of

the WT2 foundation and the temporary support had to be

removed in order to construct WT2.

4 Back Analysis of the Initial Slope Failure

The estimation of shear strength parameters along the

sliding surfaces is quite difficult in slope engineering

(Sonmez et al. 1998). The limit equilibrium back analysis

of a failed slope is one of the most reliable approaches to

determine the shear strength of slope material at the time of

failure (Sancio 1981; US Army Corps of Engineers 2003;

Topal and Akin 2009). The shear strength parameters

obtained by the back analysis of slopes are accepted as

being more consistent than those obtained by laboratory or

in situ testing during remedial measure design (Popescu

and Schaefer 2008). In conventional back analysis, the

internal friction angle or cohesion is assumed in order to

calculate the other parameter, considering a factor of safety

of 1.0. Although back analysis based on linear failure cri-

terion is mostly applied in soil slopes, the same procedure

can be followed on very weak rock mass, which is trans-

formed into a soil-like material as a consequence of

chemical weathering or alteration (Cai et al. 2007; Sha-

rifzadeh et al. 2010). On the other hand, in recent years, the

geotechnical characterization of homogeneous and isotro-

pic rock masses has mostly been performed using theFig. 3 Gray graphitic shale (a), the alternation of sandstone and shale

(b), a close-up view of heavily jointed slope material (c)

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 361

123

geological strength index (GSI) system (Morales et al.

2004; Marinos et al. 2006). The GSI system proposed by

Hoek et al. (1995) allows the determination of rock mass

strength and deformation parameters for both hard and

weak rock masses.

The back calculation of shear strength parameters of

sliding surfaces using the linear Mohr–Coulomb criterion is

independent from normal stress. However, the failure

envelope of a closely jointed rock mass is non-linear and is

sensitive to normal stresses (Sonmez et al. 1998; Yang and

Yin 2004). The Hoek–Brown non-linear failure criterion

(Hoek and Brown 1980; Hoek et al. 2002) has been com-

monly employed for the back analysis of slope failures in

heavily jointed rock masses (Sonmez et al. 1998; Sonmez

and Ulusay 1999; Cai et al. 2007; Sharifzadeh et al. 2010).

The shear strength parameters of a failure surface in such

rock masses can be determined for a specific normal stress

using the material constants of the Hoek–Brown failure

criterion (m and s) as a function of the rock mass rating

(RMR) system or the GSI system.

Fig. 4 The 20-m high cut slope

with two benches and first slope

failure after excavation

(modified after GDBP 2006)

Fig. 5 Separation between the

main tank and the maneuver

room after slope failure (a side

view, b overhead view)

Fig. 6 Immediate slope

remediation after the first slope

failure (modified after GDBP

2006)

362 M. Akin

123

The non-linear Hoek–Brown failure criterion for

homogeneous and isotropic rock masses is defined by the

equation below:

r01 ¼ r03 þ rci� ½mb � ðr03=rci

Þ þ s�0:5 ð1Þ

where r01 and r03 are the maximum and minimum principal

effective stresses acting upon the sliding surface, rciis the

intact rock strength, and mb and s are the material

constants, which are determined by the following

formulas in accordance with the GSI:

mb ¼ mi � exp½ðGSI� 100Þ=ð28� 14DÞ� ð2Þs ¼ exp½ðGSI� 100Þ=ð9� 3DÞ� ð3Þ

where mi is the intact material constant and D is the dis-

turbance factor of rock mass due to blasting or excavation.

The GSI value can be directly determined in the field

based on site conditions, although sampling for laboratory

testing is extremely difficult in heavily jointed sedimentary

and metamorphic rock masses such as shale, flysch, and

schist. In addition, alternative procedures may be imple-

mented in order not to overestimate the mb and s values, as

overrated input parameters may lead to unrealistic results

in the slope stability back analysis using the non-linear

approach (Unal et al. 1992; Sonmez et al. 1998). The back

analysis of failed slopes using the GSI system can be

performed with a trial and error approach following the

procedure first presented by Sonmez et al. (1998). The

calculation steps are as follows:

(a) A GSI value called GSI(s) is assessed and the material

constant s is determined using Eq. 3.

(b) The material constant mb is calculated considering the

existing slope geometry and slip surface in limit

equilibrium software using the Hoek–Brown failure

criterion given in Eq. 1, which satisfies the limit

equilibrium condition (FOS = 1.0).

(c) The calculated material constant mb in the previous

step is employed in Eq. 2 and discloses the second

GSI value, named GSI(m).

(d) The calculation step is carried out for different values

of GSI(s) to obtain a variety of GSI(s) and GSI(m) pairs.

(e) The results are presented in a GSI(s)-GSI(m) graph

and a straight line passing from the origin with an

inclination of 45� is drawn. The inserted line inter-

sects the GSI(s)-GSI(m) curve at a certain point

identifying the GSI value of the investigated rock

mass (GSIRM).

Following the back analysis, the instantaneous cohesion

and internal friction angle along the existing failure surface

can be calculated by application of the non-linear Hoek–

Brown failure criterion, considering the normal stress and

the GSIRM value.

For the actual slope failure, the shear strength parame-

ters of the sliding surface mobilized at the time of failure

were estimated by means of back analysis using the non-

linear (GSI) approach. The slope stability back analyses

were conducted using the Slide v.5.0 software (Rocscience

Inc. 2002) and the slope geometry before the failure was

considered in the analyses (Fig. 7). In addition, the slope

was kept in dry conditions in the back analysis, since no

groundwater table was observed in the field and in bore-

holes drilled after the construction of the pile wall support.

Due to the impossibility of sampling in heavily jointed

rock mass, the uniaxial compressive strength (UCS) of the

slope material was determined through block punch index

(BPI) tests (Ulusay and Hudson 2007) using thin cylin-

drical slices of rock pieces from the slope material. The

calculated BPIc (corrected BPI) was then converted to the

UCS in accordance with the equations presented by Ulusay

and Hudson (2007). In the BPI test, thin cylindrical disc-

shaped specimens prepared from cores or blocks are put

into an apparatus which is designed to fit the well-known

point load device (Ulusay et al. 2001). The specimens are

loaded and forced to break by a rectangular rigid punching

block. In this study, disc slices used in the BPI tests were

drilled from rock blocks obtained from the investigation

area. The unit weight was also determined on the same

discs. Consequently, the average unit weight and UCS of

Fig. 7 Slope geometry before

the first failure considered in the

back analyses

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 363

123

the intact slope material are 17.3 kN/m3 and 15.3 MPa,

respectively.

The GSI value of the rock mass studied was directly

determined in the field in accordance with the latest

quantitative GSI chart recommended by Sonmez and Ul-

usay (2002). More realistic GSI values can be estimated in

this chart by means of structure rating (SR) and surface

condition rating (SCR). The SR value is assigned based on

the volumetric joint count (Jv), whereas the SCR of the

discontinuities is calculated considering roughness,

weathering, and infilling parameters. The volumetric joint

count (Jv) of the slope material in the study area is around

21 with respect to in situ measurements. On the other hand,

discontinuity surfaces are generally smooth, highly

weathered, and contain soft clay infillings with a thickness

of\5 mm. Then, the SR and SCR values were found to be

27 and 4, respectively. The GSI value of the slope material

is 24, as seen in Fig. 8, indicating a blocky and disturbed

rock mass. It should be noted that the material constant mi

was selected as 4 in the back analysis with regards to the

recommended mi values for clastic rocks by Hoek et al.

(1995), because the triaxial test is almost impossible to

carry out on such rock types. Additionally, a disturbance

factor (D) value of 0.8 was employed in the back analysis

in accordance with Hoek et al. (2002), as the slope was

excavated mechanically and was subjected to a minor

disturbance due to stress relief from overburden removal.

The GSI(s)-GSI(m) graph obtained from the back anal-

ysis of the failed slope following the procedure proposed

by Sonmez et al. (1998) is illustrated in Fig. 9. The GSIRM

value of the failed slope was found to be 21, as seen in

Fig. 9. As the surface characteristics of discontinuities

were very poor and the slope material was tectonically

deformed, sheared, and jointed with a chaotic structure, the

GSIRM value of the rock mass assessed by back analysis is

reasonable and is compatible with the GSI value of 24

determined in the field.

The Hoek–Brown failure envelope of the slope-forming

rock mass was constructed using the GSI value of 24

determined in the field and the related material constants

(mb: 0.034, s: 8.5e-6, a: 0.5) calculated in accordance with

Eqs. 2 and 3 (Fig. 10). Based on the back analysis con-

sidering the pre-failure slope geometry and the location of

the sliding surface in Fig. 7, the normal and shear stresses

acting at the bottom of each slice of the observed failure

surface was calculated. These data pairs were plotted onto

the non-linear Hoek–Brown failure envelope of the

investigated rock mass, as depicted in Fig. 10. These

normal–shear stress plots mostly fall on the Hoek–Brown

failure envelope, indicating that the location of the sliding

surface, the estimated GSI value via back analysis, and the

rock mass parameters accurately represent the studied

failure.

The relationship between shear strength and normal

stress is more accurately represented by a non-linear

model. Furthermore, in the non-linear failure approach, the

shear strength parameters mobilizing on the failure surface

is normal stress-dependent. The instantaneous shear

strength parameters are obtained by the intercept and the

inclination, respectively, of the tangent to the non-linear

relationship between the shear strength and normal stress

(Hoek et al. 2002). In other words, the term ‘instantaneous’

indicates the shear strength parameters at a certain normal

stress level on the non-linear failure envelope.

Therefore, the instantaneous shear strength parameters

along the existing failure surface (ci and /i pairs) were

determined regarding the actual normal stress at the bottom

of each slice. In the analyzed slope, the variation of ci and

/i with different normal stresses is illustrated in Fig. 11.

The normal stress level on the actual sliding surface attains

a maximum value of 130 kPa. On the other hand, the

instantaneous cohesion varies between 6 and 28 kPa,

whereas the instantaneous internal friction angle changes

between 21� and 50�.

5 First Remedial Measure: Application

of the Retaining Pile Wall

After the first failure, it was planned to construct retaining

piles in front of WT1 in order to stabilize the constructed

tank. Therefore, both the safety of WT1 would be provided

and the toe buttress on the foundation of WT2 would be

removed. In addition to retaining piles, a new re-sloping was

also performed by lowering the slope angle of benches, to

decrease sliding forces. The shear strength parameters were

assessed for the pile design due to the presence of insuffi-

cient data for the slope material. Two different material

zones were distinguished during the design phase. The first

zone on the upper level of the slope was represented by

disturbed material which was affected from the first slide.

The second zone under the first subdivision was the undis-

turbed section of the rest of the slope. The slope material

parameters used for the design of the retaining piles are

presented in Table 1 (GDBP 2006). It should be kept in

mind that the shear strength values in Table 1 are not related

to the back analyses performed in this study. The new slope

model with the retaining pile wall is presented in Fig. 12.

As shown in Fig. 12, the new cut slope between the two

water storage tanks has three benches with lower inclina-

tions (54�–57�). Furthermore, 15.6-m long RC piles

(diameter 80 cm) without any anchors were proposed to

support WT1. The axial distance(s) between each pile is

1.60 m. After the preliminary design, a contiguous bored

pile wall was constructed in accordance with the submitted

support model. The slope was then re-excavated with three

364 M. Akin

123

benches. Subsequent to concrete pile wall construction and

re-sloping, the buttress at the slope toe was removed.

Additionally, a total of three 22-m deep boreholes were

drilled between WT1 and the retaining pile wall. Each hole

was cased with an inclinometer casing to monitor probable

lateral movements in the slope and the retaining pile wall.

Fig. 8 Determination of the GSI value of the slope material using the proposed chart by Sonmez and Ulusay (2002)

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 365

123

6 Second Slope Failure after Pile Wall Construction

One week after finishing the retaining pile wall construc-

tion and the removal of the toe buttress (22.04.2006), a

lateral movement along the longitudinal slope axis (parallel

to the failure direction) was noticed by inclinometer mea-

surements. Fifteen days following the first inclinometer

measurement (07.05.2006), the lateral movement attained a

maximum value of 10 mm. The inclinometer data indi-

cated that the slope in front of the retaining pile wall was

still unstable and that the slope was still moving along the

same failure surface (Fig. 13). A large-scale tension crack

was also observed on the slope benches as an obvious sign

of slope instability (Fig. 14a). In addition to the tension

crack, progressive small-scale cracks occurred adjacent to

WT1 (Fig. 14b). A granular toe buttress was once again

placed immediately to stabilize the slope. The lateral

movement in the slope was prevented after this toe buttress

application (11.05.2006), according to the inclinometer

measurements shown in Fig. 13.

The second failure after the retaining pile wall con-

struction indicated that the support was not sufficient to

provide stability for the cut slope and WT1. The slope

material facing the retaining pile wall was the only

resisting force for the bending moments on the anchorless

piles. The release of the resisting slope material after the

failure resulted in a lateral movement towards the longi-

tudinal slope axis in the retaining piles, due to the lateral

earth pressure on the backs of the piles. Besides, when the

cumulative displacement graph in Fig. 13 is observed, it

can be clearly seen that the first lateral movement started

almost from the bottom of the piles (around 12 m). Finally,

it can be concluded that the decrease of resisting forces

acting on the pile wall after the second slide caused sig-

nificant pile displacements in the contiguous retaining piles

without any anchorage.

7 Final Solution for Stability: Permanent Toe Buttress

It was of great importance to maintain the permanent sta-

bility of WT1 on the upper elevation after the first slope

failure. However, the constructed pile wall support was

unsatisfactory for slope stabilization. Therefore, an

improved solution that would result in a factor of safety

sufficient to resist additional slope movements was

implemented. Toe counterweights and buttresses are gen-

erally efficient for the mitigation of slope instability (Rowe

2001). The application of a temporary toe buttress after the

first and second slides prevented additional slope dis-

placements. Therefore, a larger buttress was constructed

Fig. 9 GSI(s)-GSI(m) graph obtained from the back analysis of the

failed slope using the non-linear approach

Fig. 10 Hoek–Brown failure envelope of the studied rock mass and

the normal–shear stress pairs acting on the observed failure surface

calculated by means of back analysis

Fig. 11 Instantaneous shear strength parameters along the existing

failure surface (ci and /i) graph

366 M. Akin

123

for efficient stabilization. However, the water storage tank

at the lower elevation (WT2) had to be shifted about 10 m

in the direction opposite to the longitudinal slope axis to

make room for the toe buttress. As previously mentioned,

the construction area was restricted by an expropriation

boundary which made the shifting quite impossible. Hence,

the expropriation boundary was officially enlarged by the

municipal council to create extra space. Consequently, the

site was expanded, which permitted the construction of

both the toe buttress and WT2.

The final slope geometry with a granular buttress is

depicted in Fig. 15. It is important to notice that the new

buttress entirely covers the slope benches and applies a

higher resisting force. No lateral displacements were

observed in the ongoing inclinometer measurements after

the installation of the new buttress. Having completed the

installation of the new buttress on the slope, the larger RC

water storage tank at the lower elevation (WT2) was con-

structed in front of the new support (Fig. 16).

8 Long-Term Stability Assessment

8.1 Estimation of Peak Ground Acceleration

The long-term stability of the final slope geometry was also

analyzed by the slope stability analysis considering the

seismic effect in this study. The project area is located in a

seismically active zone in the Western Anatolia Region. A

Table 1 Specific slope material parameters used for the retaining pile wall design (GDBP 2006)

Unit weight (cn)

(kN/m3)

Cohesion

(c) (kPa)

Internal friction

angle (/) (�)

Modulus of elasticity

(E) (MPa)

Poisson’s

ratio (t)

Zone 1 (disturbed

material)

19 24 27 200 0.35

Zone 2 (undisturbed

material)

19 35 28 500 0.35

Fig. 12 New slope model with

retaining piles and re-sloping

(modified after GDBP 2006)

Fig. 13 Cumulative displacement graph from inclinometer 1 (paral-

lel to the longitudinal slope axis) (modified after GDBP 2006)

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 367

123

significant extensional regime in this region resulted in

numerous normal faults and graben systems (Bozkurt

2001). The Manisa Fault exists in the very close vicinity of

the project area. This normal fault is about 40 km in length

and lies in the southern margin of Manisa city (Fig. 17).

Although a moment magnitude (Mw) of 5.2 was recorded in

1994 in this fault segment (Emre et al. 2005), Kayabali and

Akin (2003) and Ulusay et al. (2004) assigned values of 7.2

and 7.4, respectively, to the Gediz Graben which is 150 km

long in total and is the main tectonic feature around the

study area. Therefore, the maximum expected earthquake

with a moment magnitude of 7.4 was considered in the

long-term stability assessment in this study. Additionally,

the epicentral distance (Re) to the main segment of the

Gediz Graben is around 25 km.

The peak ground acceleration in the project area was

evaluated by two different regional attenuation relation-

ships of Ulusay et al. (2004) and Kayabali and Beyaz

(2011), given in Eqs. 4 and 5, respectively. In these

equations, PGA is the peak ground acceleration (cm/s2),

Fig. 14 Slope failure-related problems after the construction of the retaining pile wall (a tension crack on slope benches, b small-scale crack

near the water tank, c retaining pile wall and slope, d inclinometer casing between WT1 and retaining pile wall)

Fig. 15 Final slope geometry

with a granular toe buttress

(modified after GDBP 2006)

368 M. Akin

123

Fig. 16 Final slope design

(a WT1 and upper buttress,

b WT2 and lower buttress,

c complete view of WT2)

Fig. 17 Simplified map of

graben systems around the study

area (modified from Bozkurt

2001)

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 369

123

Mw is the moment magnitude, and Re is the epicentral

distance (km):

PGA ¼ 2:18e0:0218ð33:3Mw�ReÞ ð4Þ

log PGA ¼ 2:08þ ð0:0254M2wÞ þ ð�1:001 logðRe þ 1ÞÞ

ð5Þ

The attenuation relationship proposed by Ulusay et al.

(2004) resulted in a peak ground acceleration of 272 cm/s2,

whereas the relationship of Kayabali and Beyaz (2011)

resulted in a PGA of 113 cm/s2. Therefore, the maximum

peak ground acceleration (272 cm/s2) determined by

Ulusay et al. (2004) was accepted for the project area.

8.2 Determination of the Seismic Coefficient

In the seismic slope stability analysis, the determination of

the seismic load acting on the analyzed slope is of great

importance. A pseudostatic approach is mostly used in

seismic slope stability analysis, where the effects of an

earthquake are represented by constant vertical and/or

horizontal accelerations (Kramer 1996). Appropriate

pseudostatic coefficients should be selected, as the seismic

coefficient is a measure of the pseudostatic force on the

slope. However, there are no certain rules for the deter-

mination of the pseudostatic coefficient in the literature

(Kramer 1996). Hynes-Griffin and Franklin (1984) sug-

gested that appropriate pseudostatic coefficients for earth

slopes should be one-half of the peak ground acceleration.

For this reason, a maximum of 136 cm/s2 horizontal seis-

mic load (one-half of 272 cm/s2) on the analyzed slope is

taken into consideration in this study.

8.3 Long-Term Stability

Long-term stability of the analyzed rock mass should be

maintained, as the close vicinity of the project area is

surrounded by residential places and a slope failure

may lead to both economic and human loss due to a

significant overflow from water storage tanks. Therefore,

the long-term stability of the supported slope design was

investigated using the rock mass parameters determined

by back analysis. The Slide v.5.0 software (Rocscience

Inc. 2002) was employed during analysis. It should be

noted that the non-linear failure criterion was taken into

consideration. The factor of safety of the final slope

design in static conditions is 1.95. In seismic conditions,

considering a maximum of 136 cm/s2 horizontal seismic

load, the factor of safety decreases to 1.52 (Fig. 18). The

calculated safety factor is acceptable even in seismic

conditions, considering the degree of risk in the project

area.

9 Conclusions

In this paper, the repeated failure of an excavated slope in

heavily jointed shale and sandstone units with a chaotic

structure was evaluated via back analysis considering the

non-linear approach. When compared with field estima-

tions, the geological strength index (GSI) value obtained by

back analysis yields satisfactory results. Furthermore, the

estimated failure surface of the analyzed slope was verified

by comparing normal–shear stress plots versus the Hoek–

Brown failure envelope derived from the field-based GSI

value. It should be kept in mind that the shear strength

parameters are normal stress-dependent in such closely

jointed rock masses and the non-linear failure approach

gives more realistic results. Therefore, assigning specific

shear strength parameters during the design phase may lead

to excessive work and insufficient remedial measures in

such slope stability problems. Finally, the slope design

with permanent granular counterweight seems to be quite

stable in accordance with the limit equilibrium analysis

performed using the non-linear approach in this study. The

most important message derived from this case study is that

proper engineering is important to avoid failure of engi-

neering structures.

Fig. 18 Stability analysis of the

final slope supported by a

permanent granular buttress

370 M. Akin

123

Acknowledgments The author is grateful to the General Directorate

of the Bank of Provinces (GDBP) for providing the necessary tech-

nical information about the project. The author also thanks Dr. Samad

Joshani-Shirvan and Dr. Margaret Sonmez for their comments on the

use of language. The author would like to express his sincerest

gratitude to Prof. Dr. Resat Ulusay for his valuable comments and

assistance during the block punch index (BPI) tests. Akademi Soil and

the Rock Mechanics Laboratory deserve thanks for the sample

preparation before BPI testing. Thanks are due to the anonymous

reviewers for their valuable and constructive comments.

References

Anderson MG, Richards KS (1987) Slope stability: geotechnical

engineering and geomorphology. Wiley, New York

Bozkurt E (2001) Neotectonics of Turkey—a synthesis. Geodin Acta

14:3–30

Cai M, Kaiser PK, Tasaka Y, Minami M (2007) Determination of

residual strength parameters of jointed rock masses using the

GSI system. Int J Rock Mech Min Sci 44:247–265

Calderon PA, Adam JM, Paya-Zaforteza I (2009) Failure analysis and

remedial measures applied to a RC water tank. Eng Fail Anal

16:1674–1685

Emre O, Ozalp S, Dogan A, Ozaksoy V, Yildirim C, Goktas F (2005)

Active faults of Izmir and its close vicinity and their earthquake

potentials. General Directorate of Mineral Research and Explo-

ration, Ankara, Turkey, Report No. 10754 (in Turkish)

General Directorate of the Bank of Provinces (GDBP) (2006) Manisa

potable water network system construction project reports

(unpublished, in Turkish)

Hoek E, Brown ET (1980) Underground excavations in rock. The

Institute of Mining and Metallurgy, London

Hoek E, Kaiser PK, Bawden WF (1995) Support of underground

excavations in hard rock. Balkema, Rotterdam

Hoek E, Carranza-Torres CT, Corkum B (2002) Hoek–Brown failure

criterion—2002 edition. In: Proceedings of the 5th North

American rock mechanics symposium, Toronto, Canada, July

2002, vol 1, pp 267–273

Hynes-Griffin ME, Franklin AG (1984) Rationalizing the seismic

coefficient method. Miscellaneous Paper GL-84-13, U.S. Army

Corps of Engineers Waterways Experiment Station, Vicksburg,

Mississippi

Kayabali K, Akin M (2003) Seismic hazard map of Turkey using the

deterministic approach. Eng Geol 69(1–2):127–137

Kayabali K, Beyaz T (2011) Strong motion attenuation relationship

for Turkey—a different perspective. Bull Eng Geol Environ

70:467–481

Kramer SL (1996) Geotechnical earthquake engineering. Prentice

Hall, New Jersey

Lee S-G, Hencher SR (2009) The repeated failure of a cut-slope

despite continuous reassessment and remedial works. Eng Geol

107:16–41

Marinos V, Fortsakis P, Prountzopoulos G (2006) Estimation of rock

mass properties of heavily sheared flysch using data from

tunnelling construction. IAEG2006, paper number 314

Morales T, Uribe-Etxebarria G, Uriarte JA, Fernandez de Valderrama

I (2004) Geomechanical characterisation of rock masses in

Alpine regions: the Basque Arc (Basque–Cantabrian basin,

Northern Spain). Eng Geol 71:343–362

Okay AI, Altiner D (2007) A condensed Mesozoic succession north of

Izmir: a fragment of the Anatolide–Tauride platform in the

Bornova Flysch Zone. Turk J Earth Sci 16:257–279

Ozdemir A, Delikanli M (2009) A geotechnical investigation of the

retrogressive Yaka Landslide and the debris flow threatening the

town of Yaka (Isparta, SW Turkey). Nat Hazards 49:113–136

Popescu ME, Schaefer VR (2008) Landslide stabilizing piles: a

design based on the results of slope failure back analysis. In:

Chen Z, Zhang JM, Li ZK, Wu FQ, Ho K (eds) Landslides and

engineered slopes from the past to the future, vols 1 and 2. CRC

Press, Boca Raton, pp 1787–1793

Rocscience Inc. (2002) Slide version 5.0—2D limit equilibrium

slope stability analysis. Rocscience, Toronto, Ontario, Canada.

http://www.rocscience.com

Rowe RK (2001) Geotechnical and geoenvironmental engineering

handbook. Kluwer, Dordrecht

Sancio RT (1981) The use of back-calculations to obtain shear and

tensile strength of weathered rocks. In: Proceedings of the

international symposium on weak rock, Tokyo, Japan, Septem-

ber 1981, pp 647–652

Sharifzadeh M, Sharifi M, Delbari SM (2010) Back analysis of an

excavated slope failure in highly fractured rock mass: the case

study of Kargar slope failure (Iran). Environ Earth Sci

60:183–192

Sonmez H, Ulusay R (1999) Modifications to the geological strength

index (GSI) and their applicability to stability of slopes. Int J

Rock Mech Min Sci 36:743–760

Sonmez H, Ulusay R (2002) A discussion on the Hoek–Brown failure

criterion and suggested modifications to the criterion verified by

slope stability case studies. Yerbilimleri Bull Earth Sci Appl Res

Centre Hacettepe Univ 26:77–99

Sonmez H, Ulusay R, Gokceoglu C (1998) A practical procedure for

the back analysis of slope failures in closely jointed rock masses.

Int J Rock Mech Min Sci 35(2):219–233

Topal T, Akin M (2009) Geotechnical assessment of a landslide along

a natural gas pipeline for possible remediations (Karacabey-

Turkey). Environ Geol 57:611–620

Ulusay R, Hudson JA (eds) (2007) The blue book—the complete

ISRM suggested methods for rock characterization, testing and

monitoring: 1974–2006. ISRM & ISRM Turkish National

Group, Ankara, p 628. ISBN 978-975-93675-4-1

Ulusay R, Gokceoglu S, Sulukcu S (2001) Draft ISRM suggested

method for determining block punch strength index (BPI). Int J

Rock Mech Min Sci 38:1113–1119

Ulusay R, Tuncay E, Sonmez H, Gokceoglu C (2004) An attenuation

relationship based on Turkish strong motion data and iso-

acceleration map of Turkey. Eng Geol 74:265–291

Unal E, Ozkan I, Ulusay R (1992) Characterization of weak, stratified

and clay-bearing rock masses. In: Hudson JA (ed) Rock

characterization: ISRM symposium, Eurock ’92, Chester, UK,

September 1992. British Geotechnical Society, London,

pp 330–335

US Army Corps of Engineers (2003) Manual on slope stability.

Manual no. EM1110-2-1902

Wyllie DC, Mah CW (2004) Rock slope engineering: civil and

mining, 4th edn. Spon Press, Taylor & Francis Group, New York

Yang X-L, Yin J-H (2004) Slope stability analysis with nonlinear

failure criterion. J Eng Mech 130(3):267–273

Slope Stability Problems and Back Analysis in Heavily Jointed Rock Mass 371

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