Hydraulics of Sewage Treatment Plants Sec-6

Section 6 Hydraulics of Sewage Treatment Plants 6 -1

2000 Assoc.Prof R.J.Keller Manual of PracticeHydraulics of Sewerage Systems

Section 6: Hydraulics of Sewage Treatment Plants

6.1 Introduction

In the design of many sewage treatment plants, the hydraulics have frequentlybeen given scant regard. This often leads to severe operational difficulties suchas component units overflowing under peak conditions and component unitsnot coping if some are out of service for maintenance purposes.

The efficient operation of sewage treatment plants requires an understandingof the hydraulic aspects of the various flow processes occurring.

Treatment plants utilise unit operations and unit processes to achieve thedesired degree of purification. In unit operations, the treatment or removal ofcontaminants is brought about by physical forces. On the other hand, in unitprocesses, the treatment occurs through chemical and biological reactions.

This chapter does not provide full details on the design and operation of asewage treatment plant. Such details may be found in other texts.

This chapter is concerned with the hydraulic design of sewage treatmentplants. Although this is often seen as a challenging exercise, the hydraulicprinciples involved in individual units are normally reasonably basic. Thechallenge lies in understanding how individual units interact hydraulicallywith each other.

Broadly, the aims of this chapter are two-fold:

1) To identify the hydraulic principles associated with various unit operationsand processes.

2) To understand how knowledge of the hydraulics leads to improved systemdesign.

The following section presents a discussion of the hydraulic aspects in broadoutline. In later sections, the hydraulics of individual unit operations andprocesses are studied and the interaction among the various units is studied.Finally, the concept of the complete hydraulic profile is considered in somedetail.

6.2 Broad Concepts

For each unit process and unit operation, the hydraulic calculations willrequire the application of one or more of the fundamental concepts, developedin Chapter 1 of this Manual. A typical example is shown in Figure 6.1 whichshows the unit process hydraulics for a clarifier.

Calculation of the water surface elevation difference between the effluentmanhole and the clarifier would require the use of the following principles:

• Pressure flow equation and pipe fitting equation for determining the headloss in the pipe between the manhole and the clarifier.



• Side overflow weir equation for calculating the highest water surfaceelevation in the effluent launder of the clarifier.

• V-notch weir equation for calculating the head on the weir crest.

Moving further upstream from the clarifier, the engineer may need to considerthe head loss across the influent ports into the clarifier, and the head loss in theinfluent pipe.

Figure 6.1: Schematic of Hydraulics for a Typical Clarifier

For each unit process, the designer must understand how the sewage flowsthrough and what water depths are required for the process. Within eachprocess, various means are used to distribute flow, maintain a certain waterdepth, and control the flow. Such means include weir gates, valves, weirs,baffles, orifices, launders and under-drains. Each of these imposes a head losson the system and must be considered in the hydraulic calculations.

Each unit process, its respective flow devices, and interconnecting piping mustbe carefully analysed. As a consequence, the water surface elevations can becalculated and the structure elevations and pumping needs can be established.This information can be summarised and presented in the form of thehydraulic profile through the entire sewage treatment plant.

In carrying out the unit process hydraulic calculations, the designer shouldconsider the need to control and equally distribute the flow into multiple tanksor within a single tank. Wherever possible, static devices - such as distributionboxes, channels, weirs, and header pipes - are better suited than dynamicdevices. The latter include modulating ports, gates, and valves. Each requires acontrol system which has inherent disadvantages of potential failure and highmaintenance.

Each unit process has particular hydraulic characteristics that should beaddressed. In the following sections, some of the major issues for



consideration in the unit processes of sewage treatment plants are discussed.The final section deals with the development of the hydraulic profile.

6.3 Principles of Sedimentation

6.3.1 Preliminary

Sedimentation is the separation of suspended particles, heavier than thewastewater, by gravity. It is one of the most widely used unit operations insewage treatment plants.

In primary treatment, sedimentation is the main unit process and is used forgrit removal and removal of other particulate matter. It is responsible forremoving 50-70% of suspended solids. The removed suspended solids containbetween 25 and 40% of the BOD.

Following biological (secondary) treatment, sedimentation is used to removethe biological floc in the activated sludge settling basin and for solidsconcentration in sludge thickeners.

In most cases, the purpose of secondary sedimentation is to produce a clarifiedeffluent which may be directly discjarged into inland waterways. Where it isused for solids concentration, the aim is to produce a sludge which can beeasily handled and treated.

An understanding of the principles of sedimentation is necessary for theeffective design of sedimentation tanks. Within such tanks, three processesmay take place as follows:

Sedimentation, defined as the removal of particles by settling under gravity.

Clarification, which is similar to sedimentation but refers specifically to theremoval of suspended matter to give a clarified effluent.

Thickening, in which settled impurities are concentrated and compacted on thefloor of the tank and in the sludge collecting hoppers.

In this section, the different classes of sedimentation are identified. Thehydraulics involved in each is then discussed and outline analyses presented.In later sections of this chapter, the use of these concepts in design isdiscussed.

6.3.2 Classes of Sedimentation

Sedimentation is classified according to the nature of the particles to beremoved and their concentration. Individual particles may be discrete, such assand and grit; or flocculent, such as organic materials and biological solids.Particle concentrations may vary from very low, through moderate, to highconcentrations in which adjacent particles are in contact. Commonly, fourclasses are identified and these are summarised in Table 6.1.



Sedimentation Class Description Application

Class 1

(Discrete particlesettling)

Sedimentation ofparticles in suspension oflow solids concentration.Particles settleindividually withoutinteraction withneighboring particles.

Removal of grit andsand particles fromsewage.

Class 2

(Flocculent settling)

Dilute suspension ofparticles which flocculateduring the sedimentationoperation. Flocculationcauses the particles toincrease in mass andsettle at a faster rate.

Removes somesuspended solids inprimary settling unitsand in upper parts ofsecondary units.Removes chemicalflocculent in settlingtanks.

Class 3

(Hindered settling andzone settling)

In suspensions ofintermediateconcentration, inter-particle forces hinder thesettling of neighboringparticles. The mass ofparticles tends to settle asa unit with individualparticles remaining infixed positions withrespect to each other. Asolids-liquid interfacedevelops at the top of thesettling mass.

Process often occurs insecondary settling unitsused in conjunctionwith biologicaltreatment facilities

Class 4

(CompressionSettling)

The concentration ofparticles is so high that astructure is formed.Further sedimentation canonly occur throughcompaction of thestructure. Compactiontakes place through theweight of the particleswhich is continuouslyincreased bysedimentation from theover-lying liquid.

Usually occurs withinthe lower layers of asludge mass. It occursat the bottom of deepsecondary settling units.It is particularlyimportant in activated-sludge final settlingtanks where theactivated sludge mustbe thickened forrecycling to the aerationtanks.

Table 6.1: Classes of Sedimentation Phenomena



6.3.3 Class 1 Sedimentation

Because the particles are considered to settle independently of neighboringparticles, Class 1 sedimentation can be analysed with reference to a singleparticle.

The terminal velocity of a discrete particle settling in a fluid is reached whenthe drag force, associated with the motion of the particle, is equal to thesubmerged weight of the particle.

For a particle of diameter d, density ρ, falling at a terminal velocity vp, in afluid of density ρf, the submerged weight, W, is given by:

( )W g df= −ρ ρ π 3

6(6.1)

The drag force on the particle is given by:

F C d vD D p= π ρ2

2

412

(6.2)

The equilibrium condition is reached when W is equal to FD – ie:

( )ρ ρ π π ρ− =f D pg d C d v3 2

2

6 412

(6.3)

Re-arrangement of Equation (6.3) yields:

( )v gd

CpD

f=−4

3ρ ρ

ρ(6.4)

The drag coefficient, CD, is not constant but varies with Reynolds Number andparticle shape. Furthermore, the particle diameter and density are usually notknown and the particles are irregular in shape. This means that Equation (6.4)cannot normally be used in practice.

Despite this, Equation (6.4) does show that the terminal velocity, vp, isdependent on particle and fluid properties and this is of value in understandingsedimentation behaviour. Furthermore, it is known that the terminal velocity inpractice is reached very quickly. Consequently, for non-flocculent particlesand uniform fluid flow, the settling velocity is effectively constant throughoutthe settling time.

In the following, this concept is applied to settling in an ideal sedimentationtank. It is shown that this leads to identification of an important designparameter, the surface loading rate.

Three common types of sedimentation tank are shown schematically in Figure6.2. These are classified as (a) Rectangular Horizontal Flow Tanks, (b)



Circular Radial Flow Tanks, and (c) Upflow Tanks. In each, four zones maybe identified as follows:

Inlet Zone: In which momentum is dissipated.

Settling Zone: In which quiescent settling occurs as the water flows towardsthe outlet.

Outlet Zone: In which the flow converges upwards to the decanting weirs orlaunders.

Sludge Zone: In which settled material collects and is removed by sludgehoppers.

Figure 6.2: Schematics of Different Types of Settling Tank

Considering first the rectangular horizontal flow tank of Figure 6.2, it isevident that the critical particle for design purposes is that which enters thetank at point A and settles at the end of the tank at point B. This particlemoves through the tank with a horizontal velocity component of Vh and avertical component of Vp (the terminal velocity).

Noting that the effective length and height of the tank are respectively L andH, the time required for the particle to settle is given by:

t HV

LVp h

= = (6.5)

Now, V QBHh = , where B is the width of the tank. Substitution into Equation

(6.5) yields:

V QBLp = (6.6)

It is clear that BL is equal to the tank surface area, A, so that:



V QAp = (6.7)

Equation (6.7) states that the slowest moving particles which could beexpected to be completely removed in an ideal sedimentation tank would havea settling velocity of Q/A. This parameter is called the surface loading rate oroverflow rate and is a fundamental parameter governing sedimentation tankperformance.

A similar analysis may be developed for the circular radial flow tank asfollows:

With reference to Figure 6.2, the detention time is given by:

t HV

drVp r

R

R= =

1

2 (6.8)

Now, V QrHr =

2π and substitution into Equation (6.8) yields:

t HQ

rdrR

R= 2

1

2π (6.9)

Evaluation of the integral leads to:

( )t

R R HQ

=−π 2

212

(6.10)

Now, ( )π R R22

12− is equal to the surface area, A. Substituting into Equation

(6.10), and noting that t HVp

= from Equation (6.8):

V QAp = (6.11)

which is identical to Equation (6.7).

Considering now the upflow tank of Figure 6.2, it is clear that the minimumupflow velocity, Vu, is equal to Q/A. The limiting case for particle removaloccurs when Vu=Vp, from which:

V QAp = (6.12)

which is identical to Equations (6.7) and (6.12).

Ideally then, all particles with a settling velocity greater than Q/A will becompletely removed from the fluid. Additionally, however, it is evident that



for tanks type (a) and (b), particles with lower settling velocities of vp/n will beremoved in the proportion 1/n. It should be noted, however, that in an upflowtank of type (c), no particles with settling velocities less than Q/A can beremoved.


Under quiescent conditions, suspended particles exhibit a natural tendency toflocculate. The settling characteristics of flocculating sediments are differentfrom those of Class 1 sediments because the various sized particles settle atdifferent rates. As larger, faster-settling particles overtake slower settlingparticles, they may collide and flocculate, forming larger aggregates with anincreased settling velocity. Thus, the typical path followed by such flocculantparticles is curved. The situation is shown schematically in Figure 6.3.

Figure 6.3: Effect of Tank Depth on Removal of Class 1 and Class 2Particles

It is an important requirement of sedimentation tanks for flocculentsuspensions that sufficient depth is available to provide the opportunity forparticle aggregation to occur. This becomes clear through an examination ofFigure 6.3, which compares the behaviour of Class 1 and Class 2sedimentation if the tank depth is reduced.

For the tank shown with a depth of H, path ACB represents the settling pathfor a critical Class 1 sediment, and path ADB that for a flocculent particle. Forthe latter, the instantaneous settling velocity is the tangent to the curve.

Now, consider the effect of reducing the depth of the tank to H/2. The forwardvelocity will be doubled and the total time of travel through the tank will behalved. The settling path followed by the critical Class 1 sediment will now beAX1, while that of the Class 2 sediment will be AY1.



Thus, it can be seen that the critical Class 1 sediment will stil just reach thebottom of the settling zone. The Class 2 sediment, however, will not havereached the tank floor and will be drawn off in the tank effluent.

It is evident that the minimum average settling velocity for particles to beremoved is the surface loading rate. However, by comparison with Class 1sedimentation, removal of Class 2 sediments depends on the depth ordetention time provided, in addition to the surface loading rate.

Now, the detention time, t, is given by:

tQ

= Tank volume (6.13)

Then, for a rectangular tank:

t BLHQ

HQ

A= = (6.14)

Equation (6.14) demonstrates that if any two of the three parameters detentiontime, depth, and surface loading rate are given, the third is fixed.

Ideally, the effects of depth and detention time on solids concentration isobtained by examining representative samples obtained at various depths.These are not usually available, especially for new schemes, and use is madeof standard values.

Class 2 sedimentation removes a portion of the suspended solids in untreatedsewage in primary clarifiers, suspended solids in the upper portions ofsecondary clarifiers, and the chemical floc in settling tanks.


Class 3 sedimentation is associated with an increased concentration ofparticles in the suspension. A condition is eventually reached where theparticles are so close together that the velocity fields of the fluid displaced byadjacent particles overlap. Additionally, there is a net upward flow of liquiddisplaced by the settling particles, resulting in a reduced particle settlingvelocity. For this reason, Class 3 sedimentation is frequently called “hindered”settling.

Most commonly, hindered settling occurs in the extreme case where the veryhigh particle concentration causes the whole suspension to settle as a blanket.Under these conditions, several distinct zones may be observed, separated byconcentration discontinuities, and this leads to the descriptive term of “zone”settling.

Figure 6.4 shows a typical batch settling column test on an activated sludge.The slope of the settling curve represents the settling velocity of the interfacebetween the suspension and the clarified liquid.



Class 3 sedimentation frequently occurs in secondary settling clarifiers used inconjunction with biological treatment facilities. In designing such clarifiers,the major design parameter is the surface loading parameter because, if thesurface loading parameter is greater than the zone settling velocity, solids willbe carried out by the effluent.

Design applications for clarifiers where Class 3 sedimentation may occur areconsidered in a later section.

Figure 6.4: Suspension Exhibiting Hindered Settling Behaviour


Class 4 sedimentation is characterised by particle concentrations which are sohigh that adjacent particles are actually in contact with each other.Consequently, a structure is formed and further settling can only occur throughcompression of the structure.

Compression takes place through a continuous increase in the weight ofoverlying particles. These are constantly added to the structure bysedimentation from the supernatant liquid. Under the increased load, the voidspaces in the structure are gradually diminished and water is squeezed out ofthe matrix.

Class 4 sedimentation usually occurs in the lower layers of a deep sludgemass, for example in the bottom of deep secondary settling facilities, and insludge thickening facilities. It is particularly important in activated sludge finalsettling tanks where the activated sludge must be thickened for recycling to theaeration tanks.



6.4 Hydraulics of Screens

6.4.1 Preliminary

Screening of sewage is one of the oldest treatment processes. The purpose ofscreens is to remove gross pollutants from the sewage stream to protectdownstream operations and equipment from damage. For this reason, it isnormally the first unit operation used at sewage treatment plants.

Screens are classified as primary screens, secondary screens, andmicrostrainers. In this section, each type of screen is defined and its rolediscussed. The hydraulic aspects are then presented. Hydraulic designequations are then developed and their use in practice illustrated by examples.

6.4.2 Primary Screens

Primary screens are typically located at the inlet to sewage treatment plantsand also at the inlet to pumping stations. They are designed to remove coarsedebris such as rags, solids, and sticks which could cause damage by foulingpump impellers or interfering with downstream performance in sewagetreatment plants.

Primary screens are normally classified as coarse with openings of 50-150 mmor medium with openings 20-50 mm. Fine screens are typically secondaryscreens and are considered later.

There are several factors that need to be taken into account in screen design.These include the strength of the screen material and its resistance tocorrosion, the clear screen area, the maximum flow velocity through the screento prevent dislodging of screenings, the minimum velocity in the approachchannel to prevent sedimentation of suspended matter, and the head lossthrough the screen.

The analysis of a primary screen involves the determination of the head lossacross it. The head loss is primarily a function of the flow velocity and thescreen openings, but may also be dependent on bar size, bar spacing, and theangle of the screen from the vertical. Several equations have been developed,but only those most widely used are considered herein.

Figure 6.5: Schematic of Sloping Bar Screen



Figure 6.5 shows a schematic of a sloping bar screen. Application ofBernoulli’s equation yields:

h vg

h vg

lossessc1

12

2

2

2 2+ = + + (6.15)

where h1 is the upstream depth of flow

h2 is the downstream depth of flow

g is the acceleration due to gravity

v1 is the upstream velocity

vsc is the velocity through the screen

For a clean or partially blocked screen, the losses are usually incorporated intoa coefficient and Equation (6.15) is expressed as:

( )losses h h hgC

v vd

sc= = − = −∆ 1 2 22

121

2(6.16)

where Cd is a discharge coefficient with a typical value of 0.84.

Alternatively, an orifice equation may be applied in the form:

∆h vgC g

QC A

sc

d d

= = � �2

2

2

21

2(6.17)

where Q is the flow rate

A is the effective open area of the submerged screen

It should be noted that the discharge coefficient in Equation (6.17) is differentfrom that in Equation (6.16). In the latter equation, the value of Cd isdependent on screen design parameters and is supplied by the screenmanufacturer or by experimentation.

If the screens are to be manually cleaned, the effective open area should betaken as 50 % of the actual open area, representing the half-clogged condition.The head loss should be estimated under conditions of maximum flow.

If the bar screen is clean, Kirschmer’s equation may be used for estimating thehead loss as follows:

∆h Wb

hv= � �β ϑ1 33.

sin (6.18)

where β is a bar shape factor, as given in Table 6.2



W is the total transverse width of the screen

b is the total transverse clear spacing between bars

hv is the upstream velocity head =� �vg12

2

θ is the angle of the bars to the horizontal

Bar Type ββββ

Sharp-edged rectangular 2.42

Rectangular with semicircular upstream face 1.83

Circular 1.79

Rectangular with semicircular upstream anddownstream face

1.67

Tear shape 0.76

Table 6.2: Bar Shape Factor for Kirshmer’s Equation

It should be noted that Kirshmer’s equation is a general form of the standardhead loss equation:

∆h K vg

=2

2(6.19)

where v is identified as v1

K is given by K Wb

= � �β θ1 33.

sin

It should be noted that the expressions developed above are of use indetermining the minimum energy losses through screens , but are of littlevalue in determining the energy loss once material begins to accumulatebehind the screen.

Design should take into account the maximum increase in head loss likely tooccur under the conditions of maximum flow rate and minimum cleaningfrequency. It is especially important with manually raked screens thatsufficient freeboard is provided in the upstream channel to avoid the danger ofovertopping at high flows.



Example 6.1

A mechanically cleaned wastewater bar screen is constructed using 6.5 mmwide bars with a clear spacing of 5.0 cm. The wastewater flow velocity in thechannel immediately upstream of the screen will vary from 0.4 m/sec to 0.9m/sec.

Determine the design head loss for the screen at the two extremes of flow.(Assume that the discharge coefficient has a value of 0.84.)

Solution:

( )Head Loss = −12 2

212

gCv v

dsc

If v1 is given, vsc can be calculated, knowing the screen geometry.

Continuity:

v h w v h wsc sc clear1 1 1 1= ( )

wwsc clear

1

( )

= bar spacing + bar widthbar spacing

= +50 6550

.

= 1.13

∴ vsc = 1.13v1

( ) ( )∴ = −∆h v v12 9 81 084

11322

12

12

x xx

. ..

= 0.02v12

v1 = 0.4 m/sec ∆h = 3.2 mm

v1 = 0.9 m/sec ∆h = 16.2 mm

Primary screens may be manually cleaned or mechanically raked. Manuallycleaned screens are only fitted in small treatment plants, typically servicing apopulation equivalent (PE) of less than 5,000. Mechanically raked screens arerecommended for all plants servicing a PE greater than 2,000.

Figure 6.6 shows a schematic of a manually raked screen. The maximum clearspacing between bars is typically set at 25 mm, although American practice



permits spacings up to 50 mm. To facilitate cleaning, the bars are normally setat 30 – 450 from the vertical.

The screenings are manually raked on to a perforated plate where they drain,prior to removal. Cleaning must be frequent to avoid clogging. Infrequentcleaning may result in significant upstream backwater caused by he buildup ofsolids. When cleaning is carried out, the sudden release of the ponded waterleads to flow surges.

Figure 6.6: Schematic of Manually Raked Screen

A schematic of a mechanically raked bar screen is shown in Figure 6.7.Typically, the maximum clear spacing between bars is 25 mm, althoughAmerican practice permits spacings up to 38 mm. A spacing of 18 mm isconsidered satisfactory for the protection of downstream equipment.

Figure 6.7: Schematic of Mechanically Raked Bar Screen



Mechanically raked screens are normally set at between 0 and 450 from thevertical. The use of such screens leads to reduced labour costs, improved flowconditions, and improved capture of screenings. A large number of proprietaryscreens with mechanical rakes are available. Manufacturers will normallyprovide design charts to facilitate selection of the correct screen size for aparticular service.

Figure 6.8 shows a schematic of another type of screen – a drum screen.Screenings naturally fall from the screen as it rotates above the hopper. Awater spray assists in removing screenings.

Figure 6.8: Schematic of Drum Screen

The velocity in the approach channel is normally kept between about 0.3m/sec and 1 m/sec. The lower limit is designed to prevent the settling of coarsematter while the upper limit is designed to prevent the screens being carriedaway by the flow.

An example illustrating the design technique for a screen and screen chamberis presented in Example 6.2.

Example 6.2

Design a screen and screen chamber and determine its hydrauliccharacteristics for a loading of 10,000 PE. All material larger than 12 mm is tobe screened out. The screen is a bar screen with rectangular bars of 5 mmtransverse dimension.



Note: At the peak design flow, the velocity through the screen should be 0.9m/sec

The water level downstream of the screen is controlled by adownstream long-throated flume which gives a depth of 400 mm at the peakdesign flow and 175 mm at ADWF.

In particular, a.) Determine head loss across screen

b.) Determine screen chamber width

c.) Check velocities

d.) If the screen is 50 % blocked, calculate the head lossacross it.

Solution:

Estimate loads

ADWF = 225l/day/PE

Peak flow factor = 4.7 × (PE)-0.11 (PE in thousands)

Load = 10,000 PE

∴ SDWF = 2.25Ml/day

= 26l/sec

Peak flow factor = 4.7 × 10-0.11

= 3.65

∴ Peak flow = 3.65 × 26

= 95l/sec

Bar spacing = 12mm (will screen out all larger material)

Bar thickness = 5mm



If screen velocity is 0.9m/sec for peak flow, calculate v1

v v sc1 = ××

bar spacingbar spacing bar width

= ×0 91217

.

= 0.64m/sec

a.) Determine head loss

( )hgC

v vd

sc2 22

121

2= −

( )=× ×

−1

2 9 81 0 840 9 0 642

2 2

. .. .

= 0 029. m

∴ Depth upstream of screen

= 400(mm) + 0.029(m)

= 429mm

b.) Determine screen chamber width.

From continuity, required clear screen width (Wsc clear( ) ) is

Q h W vsc clear sc= × ×1 ( )

( )∴ =×

Wsc clear

0 0950 429 0 9

.. .

= 0.246m

∴ Required screen chamber width

= ×0 2461712

.

= 0.349m or 350mm

(CHECK against approach velocity)



vQ

W h11

0 0950 349 0 429

=×

=×

.. .

=0.64m/sec

c.) Check velocities

ADWF = 0.026m3/sec

Associated h2=175mm

∴ =×

v20 026

0175 0 349.

. .

= 0 426. m / sec

Now, because the flow is lower, we would expect a reduced head loss as well.

∴ The upstream depth will be less than 0.175 + 0.029 < 0.204m

∴ >×

=v10 026

0 204 0 3490 365

.. .

. m / sec

>0.3m/sec

∴ O.K.

Note: We could calculate v1 exactly, but the above argument removes theneed to do so.

d.) Head loss with screen half blocked

Energy equation:

hvg

hvg

hL112

222

2 2+ = + +

For peak flow Q = 0.095m3/sec

h2 0 4= . m

( ) ( )∴ =× ×

−h v vL sc1

2 9 81 0 84 22

12

. .

vQ

h h h11 1 10 35

0 0950124

0 766=

×=

×=

..

..



Substitute for v h v vsc1 2 2, , , in energy equation

( ) ( )h

h h h1

2

12

2

2

2

12

2

12

0 27119 6

0 40 67919 6

119 6 0 84

0 766 0 271+ = + +

×−� �

..

..

. . .. .

∴ + − =hh1

12

0 003750 4235 0

..

Solve by trial

h1 0 539= . m

∴ Head loss = 539 – 400

=139mm

∴ = =×

vQ

hsc 01240 095

0124 0 5391..

. .

=1.42m/sec

vh1

1

0 271 0 2710 539

0 503= = =. .

.. / secm

6.4.3 Secondary Screens

Secondary screens have smaller openings than primary screens and areinstalled following pumping and ahead of the grit chamber. Their purpose is toremove material such as paper, plastic, cloth, and other particles which mayaffect the treatment process downstream; and to minimise blockages in sludgehandling and treatment facilities.

Secondary screens are analysed in the same way as primary screens. The onlydifference is in the maximum clear spacing of bars. This is typically around 12mm, although openings as small as 6 mm have been used in practice.

6.4.4 Microstrainers

Microstrainers have been used to further reduce suspended solids in effluentfrom secondary clarifiers following biological treatment. They typicallycomprise very fine fabric or screen wound around a drum. They are typicallyabout 75 % submerged and rotate with wastewater flowing from inside tooutside.

Microstrainer openings are typically from 20 – 60 µm. They are successful atremoving suspended solids, but not bacteria.



The main hydraulic aspect is the determination of the head loss, which isanalysed semi-empirically. It is observed that the head loss is directlyproportional to flow rate, degree of clogging, and time; and inverselyproportional to the surface area of the strainer. These observations lead to:

dhdt

k QA

h= (6.20)

where k is a characteristic loss coefficient.

Integration of Equation (6.20) leads to:

h h ek

QA

t= 0 (6.21)

where h0 is the head loss across the clean strainer.

The United States Environmental Protection Agency surveyed a number ofmicrostrainers treating secondary effluent with solids concentrations in therange of 6 – 65 mg/L and found average removals of between 43 and 85 %.Typical design parameters are presented in Table 6.3.

Property Typical Value

Screen Mesh 20 – 25 µm

Submergence 75 % of height

Hydraulic Loading 12 – 24 m3/m2/h

Head Loss 7.5 – 15 cm

Maximum Head Loss 30 – 45 cm

Peripheral Drum Speed 4.5 m/min at head loss of 7.5 cm

40 – 45 m/min at head loss of 15 cm

Typical Drum Diameter 3 m

Table 6.3: Typical Microstrainer Design Parameters

6.5 Hydraulics of Grit Chambers

6.5.1 Preliminary

Within sewage treatment plants, grit - comprising sand, egg shells, coffeegrounds and other non-putrescible material – may cause severe problems in



pumps, sludge digestion facilities, and de-watering facilities. In addition, itmay settle out in downstream pipes and processes.

The grit removal process is carried out at an early stage of treatment becausethe grit particles cannot be broken down by biological processes and theparticles are abrasive and wear down the equipment. Because the grit materialis non-putrescible, it requires no further treatment following removal from thesewage treatment process and ultimate disposal.

It should be noted, however, that the location of grit chambers upstream of thesewage pumps at the entrance to the sewage treatment plant, would normallyinvolve placing them at a considerable depth involving substantial expense. Itis, therefor, usually more economical to pump the sewage, including the grit,to grit chambers located at a convenient position upstream of the treatmentplant units. It is recognised that the pumps may require greater maintenance asa result.

Grit chambers are designed to remove inorganic solids of size greater thanabout 2 mm. Removal is commonly effected using settlement, separation usinga vortex, or settlement in the presence of aeration. (In the latter process,aeration keeps the lighter organic particles in suspension.) There are importanthydraulic principles associated with each of these three processes.

In this section, the choice of grit removal process is first discussed. The threemain types of grit chamber are then described and the hydraulic aspects of theoperation of each are described qualitatively and, where appropriate,quantitatively. Design aspects are also discussed.

6.5.2 Choice of Grit Removal Process

The choice of grit removal process depends largely on the size of the sewagetreatment plant. For a PE less than 5,000, a horizontal flow (constant velocity)settling chamber is commonly used.

For medium-sized treatment plants, handling a PE of between 5,000 and10,000, a vortex type grit chamber is commonly used. For plants handling aPE greater than 10,000, the aerated grit chamber is often specified, althoughthe vortex type chamber may also be used.

Whichever type is used, it is vital that the unit must operate effectively overthe full range of expected flows.

Other non-hydraulic considerations include grit removal from the unit, whichmay be manual or mechanical; handling, storage, and disposal of grit; and theprovision of standby or bypass facilities.

6.5.3 Horizontal Constant Velocity Grit Chamber

The horizontal flow grit chamber is basically an open channel with a detentiontime sufficient to allow design particles to settle. Additionally, the velocity



must be sufficiently high that organic materials are scoured so that they passthrough the grit chamber for subsequent biological treatment.

The Camp-Shields equation is commonly used to estimate the scour velocityrequired to re-suspend settled organic material. This equation is expressed as:

v kgdfs

p=−

� �8 ρ ρ

ρ(6.22)

where vs is the velocity of scour

d is the particle diameter

k is an empirical constant (typically 0.04 – 0.06)

f is the Darcy-Weisbach friction factor (typically 0.02)

ρp is the particle density

ρ is the fluid density

Typically, this equation yields a required horizontal flow velocity of 0.15 – 0.3m/sec. This compares well with the Malaysian design standard of 0.2 m/sec.

The primary hydraulic design issue for the horizontal flow grit chamber is themaintenance of the constant velocity in the channel, despite large variations inthe flow rate, based on a typical diurnal flow pattern.

The problem is illustrated in the following.

Consider a rectangular channel with the flow passing over a rectangular weir.The discharge relationship for the weir is:

Q C B gHd= 23

2 (6.23)

where Cd is a discharge coefficient

B is the channel width

H is the channel depth

The derivation of Equation 6.23 is presented in Chapter 4.

Now, the horizontal velocity, vh, is related to the flow rate, Q, and channelgeometry by:

v QBH

C B gHBH

C gHhd

d= = =2

23

2 12 (6.24)

Substituting for H1

2 from Equation (6.23) yields:



v C g QC gBh d

d

= � �22

13

(6.25)

( )

( )∴ = � �

vv

QQ

h

h

max

min

max

min

13

(6.26)

Now, a typical value for the ratio of maximum to minimum flow rates is about5. Substitution of this ratio into Equation (6.26) yields a corresponding value

for the ratio of maximum to minimum velocities of 51

3 = 1.71. If 0.2 m/sec ischosen for the value of vh(min), the corresponding value for vh(max) would be0.342 m/sec, which would be unacceptably large. Accordingly, the shape ofeither the channel or the weir must be modified to maintain a satisfactoryhorizontal velocity.

Modification of Channel Shape:

The issue to be resolved is whether or not it is possible to develop a channelshape such that the horizontal velocity remains constant for all flow rates. It isassumed that the channel discharges into a rectangular control section, such asa long-throated or Parshall flume. Such a device acts as a water level controland a flow measurement device.

The analysis that follows is generally applicable to any rectangular cross-section. The analysis specifically makes use of the properties of a long-throated flume because it is widely used in practice and the analysis of theflume has been previously presented in Chapter 4.

As shown by Equation (4.39), the flow through a long-throated flume may beexpressed in the form:

Q g b Hc= � ��

��

23

23 1

32 (6.27)

where bc is the throat width

H1 is the upstream head

Differentiation of Equation (6.27) yields:

dQ gb H dHc= 23 1

12

1 (6.28)

Now, within the channel, the horizontal velocity, vh, is given by:

v QwHh =

1

(6.29)



or:

Q v wHh= 1 (6.30)

where w is the channel width

Differentiation of Equation (6.30) yields the flow through an elementalhorizontal strip of width w in the channel in the form:

dQ v wdHh= 1 (6.31)

Equating the right hand sides of Equations (6.28) and (6.31) yields:

23 1

12

1 1gb H dH v wdHc h= (6.32)

Solution of Equation (6.32) for w yields:

w g bv

Hc

h

= 23 1

12 (6.33)

or, noting that vh is constant:

w H= constant x 1

12 (6.34)

Equation (6.34) describes a parabola, indicating that a parabolic shape for thechannel cross-section will ensure a constant value of vh, regardless of flowrate.

Design Aspects:

To reduce construction costs, the parabolic shape is normally approximatedwith a trapezoid.

As a minimum, one channel and a bypass should be installed. When thenumber of channels is determined, the maximum, average, and minimumflows in an individual channel can be determined.

The system should be designed such that, when one channel is out of service,its flow is diverted to the other channels. The resulting emergency flow foreach channel is based on the maximum flow into the set of grit chambers withone out of service.

The four flows, Qemerg., Qmax, Qave., and Qmin., are used to design the shape andlength of the grit channel.

Other practical aspects are associated with the turbulence which occurs in theinlet and outlet zones of the chamber. These zones are illustratedschematically in Figure 6.9.



Turbulence occurs in the inlet zone as the flow is established. A similarphenomenon occurs in the outlet zone as the flow streamlines turn upwards.

To allow for this disturbance, a 25 – 50 % increase in the calculated settlinglength is applied.

Typical design criteria for a channel-modified horizontal grit chamber arepresented in Table 6.4.

A schematic of a typical channel-modified horizontal grit chamber ispresented in Figure 6.10.

Design Parameter Typical Values Comments

Water depth (m) 0.6 – 1.5 Dependent on channelarea and flow rate

Length (m) 3 – 25 Function of channeldepth and grit settlingvelocity

Extra for inlet andoutlet

25 – 50 % Based on theoreticallength

Detention time at peakflow (seconds)

15 – 90 Function of velocityand channel length

Horizontal velocity(m/sec.)

0.15 – 0.4 0.2 m/sec is MalaysianStandard

Table 6.4: Typical Design Criteria for Channel-Modified GritChamber

The design procedure for a channel-modified grit chamber is illustrated inExample 6.3.

Example 6.3

Design a horizontal/constant velocity grit chamber for a hydraulic load of2,000 PE. Consider only the ADWF and the peak flow.

Note: The water level within the chamber is controlled by a downstreamlong-throated flume which gives a depth of 205 mm at the peak design flowand 80 mm at ADWF.

Maximum horizontal velocity is 0.2 m/sec

Channel length > 18 x maximum water depth

Grit quantity is estimated as 0.03 m3/ML of wastewater

Grit collection channel to be cleaned out twice per week



Figure 6.9: Schematic of Settling Process in Grit Chamber

Solution

Average dry weather flow

= 225 × 2,000

= 0.45 ML/day

= 5.2l/sec

Peak flow factor = × −4 7 2 0 11. .

= 4.35

∴ Peak flow = 4.35 × 5.2

= 23 l/sec

Flow control gives depth of 205mm at peak flow

80mm at ADWF

(Consistent with long-throated flume of throat width 133mm)



Calculate cross-sectional areas

ADWF: Area = 0.0052

0.2

= 0 026 2. m

Peak: Area = 0.0230.2

= 0115. m2

Surface widths at each flow are now calculated

Refer to Equations (6.27) and (6.33).

Q gw yt= � �23

32 3

2 (6.27)

& w gwv

yt

h=

23

12 (6.33)

Transposing Eq. (6.33)

wwv

g yt

h=23

12

Substitute in Eq. (6.27)

Q wyvh=23

∴ Cross-sectional area

=23

wy

∴ At average dry weather flow

Surface width = ×Ay

32

=×

×0 026 32 0 08.

.

= 0 49. m



At Peak Flow

Surface width =×

×0115 32 0 205

..

=0.84m

Length of chamber:

> 18 × max. depth

> 18 × 0.205

Use 3.7m

Grit quantity:

Based on average DWF

Grit quantity = 0.45 × 0.03

= 0.014m3/day

∴ At twice weekly cleanout, grit accumulation

= ×0 014 4. ~

= 0 056. m3

∴ Required cross-sectional area of grit collection channel

=0 056

37..

= 0 015. m2

Use grit collection channel 150mm wide × 110mm deep

(gives some margin)

Allow for freeboard (say, 200mm)

Parabolic section to be approximated by trapezoid



Figure 6.10: Schematic of Channel-modified Horizontal Constant VelocityGrit Chamber

Modification of Downstream Control Weir:

For a rectangular grit chamber, the flow rate is given by:

Q v Byh= (6.35)

where B is the chamber width

y is the flow depth in the chamber

The form of Equation (6.35) indicates that for vh to be constant, regardless offlow rate, the flow rate should be linearly proportional to the depth, y. Thismay be assured by using a downstream control weir characterised by a linearrelationship between flow rate and head on the weir crest.

Such a weir is the Sutro weir which is described and analysed in Chapter 4 ofthis Manual. For details and a worked example, refer to Section 4.4.4 andExample 4.5.

6.5.4 Vortex Grit Chamber

A schematic of a typical vortex grit chamber is shown in Figure 6.11.

With reference to this figure, grit-laden flow enters the unit tangentially at thetop. The resulting spiral flow pattern tends to lift the lighter organic particleswhile the mechanically induced vortex captures grit at the centre. The grit isthen removed by air-lift or through a hopper. It should be noted that the grit



sump has a tendency to become compacted and clog. Sometimes provision ismade for the use of high-pressure agitation water or air to clear the sump.

Figure 6.11: Schematic of Typical Vortex Grit Chambers (a) PISTAUnit (b) Teacup Unit

The adjustable rotating paddles maintain the proper circulation within the unitfor all flows. However, attention should be paid to the tendency for thesepaddles to collect rags.

Vortex grit chambers are highly energy-efficient. The head loss across the unitis minimal when operating correctly and unclogged. American practiceindicates a value of 6 mm, although an allowance of 100 mm is recommended.



Vortex grit chambers have the great advantage that they are very compact.Their design is usually proprietary so that manufacturers will usually producea suitable unit to accommodate stated performance specifications.Manufacturers’ specifications will provide information on the maximum waterdepth within the chamber.

6.5.5 Aerated Grit Chamber

Aerated grit chambers are commonly used in medium to large sewagetreatment plants. The introduction of air through a diffuser, located on one sideof the tank, induces a spiral flow pattern in the sewage as it moves through thetank, as shown in Figure 6.12. Correct positioning of the tank inlet and outletdirects the flow perpendicular to the spiral roll pattern. Inlet and outlet bafflesare normally installed to dissipate energy and minimise short-circuiting. Headloss across the chamber is minimal.

Figure 6.12: Helicoidal Flow Pattern in an Aerated Grit Chamber

The roll velocity is set so that it is sufficient to maintain lighter organicparticles in suspension while allowing heavier grit particles to settle. Becauseconditions change with flow rate, the air supply is adjustable to provide theoptimum roll velocity.

A further advantage of the introduction of air is that the sewage is freshened,leading to a notable reduction in odour. If desired, the chamber can be used for



chemical addition, mixing, and/or flocculation ahead of primary treatment.Grease removal may be achieved with a skimmer.

If correctly designed, an aerated grit chamber with a minimum hydraulicdetention time of 3 minutes will capture about 95% of grit larger than 0.2 mmwhen operating at its peak flow. The usual range of design specifications isgiven in Table 6.5.

The design of an aerated grit chamber is illustrated in Example 6.4.

Example 6.4

Design an aerated grit chamber for a hydraulic load of 20,000 PE.

Note: The minimum detention time at peak flow is 3 minutes

The width to depth ratio is 2:1

The length to width ratio is 2:1

Grit quantity is estimated as 0.03 m3/ML of wastewater

The aeration requirement is 10 litres/sec/m length of tank

Design Parameter Range of Values Comments

Depth 2 – 5 m Varies widely

Length 8 – 20 m

Width 2.5 – 7 m

Width:Depth Ratio 1:1 – 5:1 2:1 typical

Length:Width Ratio 3:1 – 5:1

Minimum DetentionTime

2 – 5 minutes 3 minutes typical

Air Supply 0.25 – 0.75 m3/min/m 0.45 m3/min/m typical

Diffuser Distancefrom Bottom

0.6 – 1.0 m

Transverse RollVelocity

0.6 – 0.75 m/sec

Table 6.5: Typical Design Specifications for an Aerated Grit Chamber



Solution

Average DWF = 20,000 × 225l/day= 4.500m3/day=52l/sec

Peaking factor = 4.7 × 20-0.11

= 3.38

∴ Peak flow = 52 × 3.38= 176l/sec

Grit chamber volume:Minimum detention time at peak flow = 3minutes

= 3 × 60 = 180seconds

∴ Required volume= 0.176 × 180= 31.7 = 32m3

WD

LW

= =2 2,

∴ Volume = D × W × L = 32

W = 2D, L = 2W = 4D

∴ D × 2D × 4D = 32

Dimensions ∴ D = 1.6m∴ W = 3.2m L = 6.4m

Aeration requirement

10l/sec/m length= 10 × 6.4 = 64l/sec

Grit quantity

Based on average flow rate,= 4.5ML/day × 0.03m3/ML= 135l/day

Note: Means should be provided to vary the air flow rate to control gritremoval rate and grit cleanliness.



6.6 Hydraulics of Clarifiers

6.6.1 Preliminary

Clarifiers are essentially sedimentation tanks and are used as a part of bothprimary treatment and secondary treatment processes. They may berectangular, square, or circular in shape.

A schematic of a typical circular clarifier has been presented in Figure 6.1.The flow enters at the centre of the tank and settlement takes place as the flowmoves outwards and rises. The effluent is collected in a channel or launder,which then conveys the flow to an exit channel or pipe.

This section emphasises the hydraulic aspects of the design of clarifiers.Design guidelines are first presented and the basic design procedure isreviewed. The important procedure for the design of the launder is thendiscussed. Finally, a design example is presented to aid understanding.

6.6.2 Design Guidelines

Design guidelines for primary and secondary clarifiers vary significantly fromcountry to country. Typical guidelines from American practice are presentedin Table 6.6.

Parameter Value

Primary Clarifiers

Surface loading rate For average dry weather flow For peak flow conditions

32 - 49 m3/m2/day49 - 122 m3/m2/day

Sidewater depth 2.1 – 5 m

Weir loading rate 125 – 500 m3/m/day

Secondary Clarifiers

Surface loading rate For average dry weather flow For peak flow conditions

16 – 29 m3/m2/day41 - 65 m3/m2/day

Sidewater depth 3.0 – 5.5 m

Floor slope Nearly flat to 1:12

Maximum diameter 46 m

Table 6.6: Typical Design Guidelines for Circular Clarifiers



Primary clarifiers are designed more conservatively if sedimentation is theonly treatment and if activated sludge is being returned to the primary clarifier.Rectangular clarifiers are generally designed under the same criteria ascircular clarifiers. Typical length to width ratios for rectangular primaryclarifiers range from 3:1 to 5:1, although many existing tanks are characterisedby ratios of between 1.5:1 and 15:1.

A well designed and operated primary clarifier should be capable of removingbetween 50 and 65% of the influent total suspended solids.

6.6.3 General Design Principles

Clarifiers are designed to remove the maximum amount of settleable solidsquickly and economically. The design objective is to provide sufficient timeunder quiescent conditions for maximum settling.

If all solids were discrete particles of uniform size, density, and shape theremoval efficiency of the tank would be dependent on the surface loading rateonly as discussed in Section 6.3. It was also shown in Section 6.3 that thedepth of the tank would have little influence on the removal efficiencyprovided horizontal velocities were maintained below the scouring velocity.

However, the solids are not of a regular character and the conditions underwhich they are present range from total dispersion to complete flocculation. Inpractice, the bulk of the finely divided solids reaching primary sedimentationtanks are incompletely flocculated and are susceptible to further flocculation.

Flocculation is aided by the eddying motion of the fluid within the clarifier. Itproceeds through the coalescence of fine particles at a rate that is a function oftheir concentration and of their natural ability to coalesce upon collision. Thus,the longer the process continues, the more complete the coalescence becomes.For this reason, the detention time within the clarifier is a consideration in thedesign process.

It should be noted, however, that the mechanics of flocculation are such that,as the time of sedimentation incless and less coalescence of the remainingparticles occurs. Accordingly, from the point of view of settling, there is apractical limit on the effective detention time of the sewage.

Primary and secondary clarifiers are normally designed to provide a detentiontime of between 1.5 and 2.5 hours, based on the average flow rate. It is notedthat the design criteria for Malaysian systems incorporate a time of 2 hoursbased on the peak flow rate.

Recalling the discussion of Section 6.3.4, the detention time is given by theequation:

t HQ

A= (6.14)

where Q/A is the surface loading rate.



Schematics of rectangular (horizontal flow), circular (radial flow), and square(upflow) clarifiers are presented in Figures 6.13, 6.14, and 6.15.

Figure 6.13: Schematic of Rectangular Sedimentation Tank

Figure 6.14: Schematic of Circular Clarifier

Rectangular tanks are commonly used for primary sedimentation. Theyoccupy less space than circular tanks and can be economically built side byside with common walls.

Circular tanks require careful design of the inlet stilling well to achieve astable radial flow pattern without causing excessive turbulence in the vicinityof the central sludge hopper. Inlet design is considered in subsequentparagraphs.

Upflow tanks typically have deep hopper bottoms and are common in smalltreatment plants. Their primary advantage is that sludge removal is carried outentirely by gravity. The steeply sloping sides – typically 600 – concentrate thesludge at the bottom of the hopper. A significant disadvantage is that hydraulicoverloading may cause major problems because any particles with a settlingvelocity less than the surface loading rate will not be removed, insteadescaping with the effluent.



Figure 6.15: Schematic of Upflow Clarifier

The primary design parameters are the surface loading rate and the detentiontime, both of which are normally specified in local design criteria. Followingthe specification of these parameters, the dimensioning of the tank thenproceeds as follows:

Tank Surface Area, A QQ

A= (6.36)

Tank length or Diameter, or L D A= α (6.37)

where α = L4

for rectangular tanks

απ

= 4 for circular tanks

The forward velocity is also an important aspect of the design of rectangulartanks. If this is excessive, scouring and re-suspension of the sludge will result.

The forward velocity is given by:

v QWHh = (6.38)

Incorporating Equation (6.14) for the detention time,

v Lth = (6.39)



It is evident from Equations (6.38) and (6.39) that the forward velocityinfluences the choice of length to width ratio. The maximum forward velocityto avoid the risk of scouring settled sludge is 10 to 15 mm/sec, indicating thatthe ratio of length to width should preferably be about 3:1.

Values of L/W in practice range between 3 and 6. The Malaysian DraftGuidelines specify a value of 3.

Design curves to aid in the determination of the tank geometry have beenpresented by Barnes (1981) and should be consulted for further information.

Weir Loading Rate

The weir loading rate is defined as Q/Lw where Lw is the length of the outletweir. If this value is too high, the approach current generated by the weir willextend upstream into the settling zone, creating a potential disruption of theflow pattern. A weir loading rate of between 100 and 200 m3/m/day istypically specified.

Achieving this value is a particular problem for rectangular tanks which isusually overcome by utilising multiple suspended weir troughs.

In circular tanks, the weir loading rate associated with a perimeter weir isnormally satisfactory at high flows. At low flows, however, difficulties mayarise from a weir loading rate which is too small because the consequent verysmall flow depths over the weir make the tank flow pattern very sensitive toerrors in weir levelling. This problem may be overcome by constructing theperimeter weir as a saw-tooth weir – or multiple V-notch – to increase theflow depth.

The issues of surface loading rate, detention time, and weir loading rate areillustrated by Examples 6.5 and 6.6.

Example 6.5

Two primary clarifiers are 26 m in diameter with a 2.1 m side water depth.Single effluent weirs are located on the peripheries of the tanks. For awastewater flow of 26,000 m3/day, calculate:

a.) The surface loading rate

b.) The detention time

c.) The weir loading rate

Solution

Surface area of each clarifier = =×π πD2 2

426

4

= 530 2m



∴ Total surface area = 530 × 2

= 1,060m2

∴ Total volume = 1,060 × 2.1

= 2,230m3

a.) Surface loading rate =QA

=26 0001 060

,,

= 24 5. m / m / day3 2

b.) Detention time =Volume

Flow rate

= ×2 23026 000

24,,

= 2 06. hours

c.) Weir loading rate =flow rate

weir length

=×

26 0002

,πD

=× ×26 000

2 26,π

= 159m / m / day3

Example 6.6

Determine the size of two identical circular final clarifiers for an activatedsludge system with a design flow of 20,000 m3/day, and a peak hourly flow of32,000 m3/day.

Note: The maximum surface loading rate is 33 m3/m2/day at design flow and66 m3/m2/day at peak flow.

Minimum detention time at design flow is 2 hours

Maximum weir loading rate at design flow is 125 m3/m/day



Solution

At design flow, surface area required for each tank

=×

=20,000m / day

2 33m / m / day303m

3

3 22

Check peak overflow rate

=×

32 0002 303

,

= 53m / m / day3 2

< 66m / m / day3 2 (OK)

Tank diameter

πD2

4303=

∴ =×

� �D303 4

12

π

= 19 6. m

Detention time =Tank volume

Flow rate

∴×

>Area Depth

Flow rate hours2

∴×

×Depth >

2 10,000303 24

> 2 75. m

Make depth 3.5m

(Recommended for tank diameter > 15m, depth should be 3.4m).

This will also give a reasonable detention time at peak flow rate.

Weir loading rate

=flow rate

weir length

For single sided weir, weir loading rate/tank



=×

10 00019 6

,.π

= 162m / m / day3

>125m / m / day3 (No good)

∴ Use an inboard weir channel with entry on both sides.

Set weir channel at a diameter of 18m.

∴ Weir loading rate

=× ×10 000

2 18,π

= 88m / m / day3

<125m / m / day3 (OK)

Two tanks,

Diameter 19.6m (20m?)

Depth 3.5m

Inboard weir set on diameter of 18m.

Tank Inlets

Sedimentation tank inlets must be designed to distribute the flow as uniformlyas possible so that the best possible flow pattern is maintained. The influent jethas a high amount of kinetic energy that must be dissipated.

For rectangular tanks, various baffled inlet arrangements have been usedwhich are effective for energy dissipation and flow distribution. Typicalarrangements are shown schematically in Figure 6.16.

With circular tanks, the radial flow from the inlet is inherently less stable thanthe horizontal flow in a rectangular basin. Careful design is needed to achievea stable radial flow pattern. Typical arrangements are shown in Figure 6.17 for(a) side feed, (b) vertical pipe feed, and (c) slotted vertical pipe feed. In allcases, the primary design principles are that energy must be dissipated and theflow distribution must be uniform.



Figure 6.16: Schematics of Typical Sedimentation Tank Inlets

Figure 6.17: Centre-feed Inlets for Circular Clarifiers: (a) Side Feed, (b)Vertical Pipe Feed, (c) Slotted Vertical Pipe Feed



Effluent Launder Design

Rising wastewater in a clarifier flows over a weir into a channel or launderwhich, in turn, conveys the collected effluent to the exit channel. Flow in thelaunder is classified as spatially varied because the flow rate increases withdistance along the launder. This characteristic requires the use of themomentum equation for its analysis, rather than the energy equation.

The basic flow condition is illustrated schematically in Figure 6.18 whichshows the flow spilling over the multiple V-notch weir into the launder. A fullmomentum analysis, including the effects of friction, has been presented byDroste (1997). A simplified approach is usually adequate and is presentedherein.

Figure 6.18: Definition Sketch for Flow in a Launder

The first issue is the size of V-notch weir required. The individual V-notchesare typically set out with a centre to centre spacing of between 150 and 300mm. With the number of V-notches consequently established, the flowthrough each can be determined from:

Q QNperV notch− = (6.40)

where N is the number of V-notches.

The maximum height, H, over the weir is then determined from the standardV-notch weir equation – refer to Chapter 4.4.3, Equation (4.24):

Q C g HperV notch d− = 815

22

52tanθ (6.41)

The discharge coefficient, Cd, is a function of the notch angle, θ. For θ = 900,Cd has a value of 0.58.



The head over the weir, calculated from Equation (6.41), should be increasedby a safety factor of 15%.

The next stage in the hydraulic design is to determine the maximum depth inthe launder. First, the critical depth at the discharge point of the launder iscalculated from:

( )yqLb gc = �

��

2

2

13

4(6.42)

where q QL

= and L is the length of the weir (circumference of the tank)

b is the width of the launder

The depth at the upstream end of the launder is then calculated from:

H y q xgb yc

c

= +� �2

2 2

2

122 (6.43)

where x L=2

for a circular basin.

The depth, H, calculated from Equation (6.43) should be increased by a factorof safety of 50% to allow for friction loss, freeboard, and a free fall allowance.

The derivation of Equations (6.42) and (6.43) has been presented by Droste(1997), in which further refinement is provided by including friction loss. Thisrefinement would enable the full longitudinal profile in the launder to becalculated.

It should be noted that such a refined design is rarely justified because designpractice usually ensures that the launder is hydraulically over-designed.

The design of a launder is illustrated by Example 6.7

Example 6.7

Design the overflow weirs and launders (collection channels.) for the clarifiersof Example 6.6.

Note: The critical condition is when the peak flow occurs with one clarifierout of service. The launder must be able to cope with the corresponding flow.

Solution

Weir design

One clarifier must handle peak flow.



∴ Peak weir loading rate

=× ×

32,000m / day2 18m

3

π

where: 2 represents the inflow on both sides

18 represents the diameter

= 283m / m / day3

Assume that weir comprises V-notches with spacing of 25cm centre to centre.

(This may need adjusting)

∴ Total number of V-notches

= ××

20 25

π D.

Take D as 18m, even though it will be less for the inner ring and more for theouter.

∴ Total number of V-notches

= ××

218

0 25π

.

= 452

∴ Flow per notch

=×

×32 000

3 600 241

142,

,

= 0 00082. / secm3

Now, for each V-notch, notch angle is 900 and Cd=0.58.

Q C g Hd=8

152

25

2tanθ

∴ =×

×� �H

15 0 000828 0 58 19 6

25.

. .

= 0.051m



A safety factor of 15% is normally appropriate.

∴ Allow for water depth over notch of 1.15 × 0.051

= 0.059m

= 60mm

∴ Width of V-notch at the top = 60mm ×2 = 120mm.

∴ Weir design as follows:

Launder design

Q is discharge/unit length of launder

= weir loading rate × 2 (because launder is fed from both sides)

∴ =×

×q283

3 600 242

,m / m / sec3

= 0 0066. m / m / sec3

Assume a launder width

Try 500mm

Calculate depth at launder discharge point

( )y

qLb gc = �

��

2

2

13

4

( )

=× ×

× ×� �0 0066 18

4 0 5 9 81

2

2

13.

. .π

=0.243m



Calculate maximum depth in launder at upstream end

H yq xgb y

= +� �22 2

2

0 52

.

(Note: xD

=π2

)

∴ = +× ×

×� �

× ×

�

�

��

��

H 0 2432 0 0066

182

9 81 0 5 0 2432

22

2

12

..

. . .

π

=0.419m

Increase this depth by 50% to allow for friction loss in the launder, freeboard,and free-fall allowance.

∴ Total depth to be provided in launder

= 0.419 × 1.5

= 0.629, say 0.65m

∴ Launder depth below vertex of V-notch weirs

= 0.65m

Launder width = 0.50m

6.7 Sludge Hydraulics

6.7.1 Preliminary

Sludge produced in sewage treatment plants must be conveyed from one plantpoint to another. The conditions of the sludge range from the consistency ofwater or scum to a thick sludge. It may also be necessary to pump sludge off-site for long distances for treatment and disposal. For each type of sludge andpumping application, a different type of pump may be needed.

The primary issues of concern are the type of pump to use, the computation ofhead loss in pipes carrying sludge, and other practical hydraulic aspects. Theseissues are examined in this section.

Types of pumps are briefly discussed. Simplified computations, suitable forshort lengths of pipe, are then presented. The application of rheology to head



loss computations for long-pipe calculations is then presented. Finally,practical aspects of sludge piping are briefly covered.

It should be noted, in particular, that details of sludge processes are notcovered because they are outside the scope of hydraulics concern. Specialisttexts should be consulted for these details.

6.7.2 Sludge Pumping

Although specifying a single type of pump to handle all sludges within atreatment plant is an attractive idea, the wide range of conditions imposed onsuch service normally exceeds the capabilities of a single type of pump.Fortunately, many types of pump are available to the design engineer.

Types of pumps most frequently used to convey sludge include the plunger,progressive cavity, centrifugal, torque flow, diaphragm, high-pressure piston,rotary lobe, and screw lift pumps. Specialist literature should be consulted fordetails on each. The application of different types of pump is summarised inTable 6.7.

Commonly, centrifugal pumps of non-clog design are used. Problems arise,however, over choosing the most appropriate size. These problems occurbecause, at any given speed, centrifugal pumps operate well only if thepumping head is within a narrow range. Because of the variable nature ofsludge, pumping heads may vary significantly.

Selected centrifugal pumps must have sufficient clearance to pass the solidswithout clogging, but have a small enough capacity to avoid pumping a sludgediluted by large quantities of sewage overlying the sludge blanket. It isimpractical to throttle the discharge to reduce the capacity because of frequentstoppages. For this reason, it is essential that these pumps be equipped withvariable speed drives. Where the application involves high pressure, multiplepumps may be used and connected in series.

Usually the consistency of untreated primary sludge changes during pumpingwith the most concentrated sludge being pumped first. Later, the pump musthandle a dilute sludge which has essentially the same hydraulic characteristicsas water.

This change in characteristics causes a centrifugal pump to operate farther outon its characteristic curve. It is necessary that the pump motor is sized for theadditional load and that a variable speed drive is used to reduce the flow underthese conditions. It should be noted that if the pump motor is not sized for themaximum load when pumping water at top speed, it will go on overload or bedamaged if overload devices do not function or are set too high.

In determining the operating speeds and motor power required for acentrifugal pump handling sludge, it is important that system curves bedetermined for the densest sludge anticipated, the average conditions, andwater. These system curves should be plotted on the pump characteristic crvesfor a range of available speeds.



Principle Common Types Typical Applications

Kinetic(rotodynamic)pumps

Nonclog mixed-flowpump,

Recessed-impellerpump (vortex pump,torque-flow pump)

Screw centrifugalpump

Grinder pump

Grit slurry, incinerator ash slurry

Unthickened primary sludge

Return activated sludge

Waste activated sludges fromattached-growth biologicalprocesses

Circulation of anaerobic digestor

Drainage, filtrate, and centrate

Dredges on sludge lagoons

Positive-displacementpumps

Plunger pump

Progressing cavitypump

Air-operateddiaphragm pump

Rotary lobe pump

Pneumatic ejector

Peristaltic pump

Reciprocating piston

Waste activated sludge

Thickened sludges (all types)

Unthickened primary sludge

Feed to dewatering mahines

Unthickened secondary sludges

Dewatered cakes

Other Air lift pump

Archimedes screwpump

Return activated sludge

Table 6.7: Sludge Pump Applications by Principle

The intersection of the pump curves with the system curves at the desiredcapacity yields the maximum and minimum speeds required for a particularpump. The intersection of the maximum speed pump curve with the systemcurve for water permits the determination of the power required. For thedetermination of hours of operation, average speed, and power costs, theintersection of the pump curve with he system curve for average conditions isappropriate.



6.7.3 Head Loss Determination

It is clear that the procedures in the previous section require an estimate of thehead loss in the pumping lines. The head loss depends on the rheology (flowproperties) of the sludge, the pipe diameter, and the flow velocity. It is known,further, that head osses increase with increased solids content, increasedvolatile content, and reduced temperatures. It is also known that, when theproduct of the percentage of volatile matter and the percentage of solidsexceeds 600, difficulties in pumping sludge are often experienced.

Dilute sludges such as unconcentrated activated and trickling filter sludgesbehave in a very similar manner to water. They are classified as “Newtonian”fluids. As such, the pressure drop is proportional to the velocity and theviscosity under laminar conditions, and to the square of the velocity underturbulent conditions. The head loss in pumping unconcentrated sludges may bebetween 10 and 25% greater than for water.

Concentrated sludges, however, are non-Newtonian fluids. The pressure dropunder laminar conditions is not proportional to the velocity and the viscosity isnot a constant. Primary, digested, and concentrated sludges at low velocity arecharacterised by a plastic phenomenon whereby a definite pressure is requiredto overcome resistance and start the flow. The resistance then increasesapproximately with the velocity up to a velocity of about 1.1 m/sec, definingthe upper limit of the laminar flow regime. Above about 1.4 m/sec, the flowmay be considered to be turbulent.

Within the turbulent range, the head losses for well-digested sludge may betwo to three times greater than for comparable water velocities. For primaryand concentrated sludges, the losses may be substantially greater.

Two approaches for calculating head losses are considered in the following. Asimplified approach is considered first, which is particularly suitable for shortpipe lines. A more complex method is then discussed which uses the sludgerheology and is suited to head loss calculations in long pipe lines.

Simplified Approach

The simplified approach is used to compute head losses in short pipe lines.The accuracy is adequate, especially for solids concentrations less than 3% byweight.

Firstly, the head loss of water at the same flow rate is determined, using anyone of the Darcy-Weisbach, Hazen-Williams, or Manning equations. Thishead loss is then multiplied by a factor, k, obtained from empirical curves for agiven solids content and sludge type, or for a given velocity and solids content.

The first method is suggested when the pipe velocity is greater than 0.8 m/sec,thixotropic behaviour is not considered, and the pipe is not obstructed bygrease or other material. Figure 6.19 presents the multiplication factor, k, as afunction of solids concentration for digested sludge and for untreated primaryand concentrated sludges respectively.



Figure 6.19: Head Loss Multiplication Factor for Different Sludge Typeand Concentration

The second method is less restrictive in its application and involves only thepipe velocity and solids concentration in determining the multiplication factor.Figure 6.20 presents the corresponding relationship for k.

Application of Rheology to Head Loss Computations

Where sludge must be pumped over long distances, the accuracy of theestimates of head loss becomes more important because of their increasedimpact on the design of pumping needs. For this reason, the head losscomputations should take account of the rheological properties of the sludge.

In the discussion following, a method is described which uses similar conceptsto the Darcy-Weisbach method, but with modifications to allow for sludgeproperties.

Sludge behaves like a Bingham plastic – ie it exhibits a linear relationshipbetween shear stress and flow only after flow begins. A Bingham plastic isdescribed by two constants, the yield stress, sy, and the coefficient of rigidity,η. Typical ranges of values for these two constants are presented in Figures6.21 and 6.22 respectively. It should be noted, however, that published dataare highly variable. If considered important, pilot studies should be undertakento determine the rheological data for specific applications.



Figure 6.20: Head Loss Multiplication Factor for Different Pipe LineVelocities and Concentrations



Figure 6.21: Range of Design Values for Yield Stress as a Function ofPercentage Sludge Solids

Figure 6.22: Range of Design Values for Coefficient of Rigidity as aFunction of Percentage Sludge Solids



Following the determination of the yield stress and the coefficient of rigidity,two dimensionless numbers are used to determine the pressure drop asfollows:

Reynolds Number

Re = ρηVD (6.44)

where ρ is the density of the sludge (kg/m3)

V is the average velocity in the pipe (m/sec)

D is the pipe diameter (m)

Hedstrom Number

He =D sy

2

2

ρη

(6.45)

The friction factor for the pipe-sludge system is then determined using thegraph in Figure 6.23.

Figure 6.23: Friction Factor for Sludge Analysed as a Bingham Plastic

The pressure drop in the pipe is then calculated from:

∆p f LVD

= 2 2ρ (6.46)

Equation (6.46) can be readily shown to be a form of the Darcy-Weisbachequation.



The equations and graphs presented above apply to the entire range of laminarand turbulent flows. It should be noted, however, that Figure 6.23 does notinclude any allowance for pipe roughness.

To allow for pipe roughness, it is recommended that, in addition to the aboveprocedure, the pressure drop should be calculated using a standard procedurefor water. If this process gives a higher pressure drop than that given byEquation (6.46), roughness is dominant and the pressure drop given by thewater formula will provide a reasonably accurate estimate of pressure loss.However, where use of the water formula is indicated, for worst case designconditions, a safety factor of 1.5 is recommended.

The procedure for calculating head loss, including the sludge rheology, isillustrated in Example 6.8.

Example 6.8

A pipeline of length 1,000 m and diameter 200 mm conveys untreated (raw)sludge at an average flow rate of 40 l/sec.

Calculate the head loss in the pipeline.

Analysis of the sludge indicated the following rheological data:

Yield stress: sy = 1.1 N/m2

Coefficient of rigidity: η = 0.035 kg/m/sec

Specific gravity: S.G. = 1.01

Solution

Pipe flow velocity

VQA

=

=×

0 0400 2

4

2.

.π

=1.27m/sec

Sludge density= 1,000 × 1.01= 1,010kg/m3

Reynolds Number

Re =ρ

µVD

=× ×1 010 127 0 20 035

, . ..

=7.33 × 103



Hedstrom Number

He =2D syρµ 2

=× ×0 2 11 1 0100 035

2

2

. . ,.

= ×363 104.

From friction factor diagramf = 0.007

∴ =∆pf LV

D2 2ρ

=× × × ×2 0 007 1 010 1 000 127

0 2

2. , , ..

=114.03kN/m2

∴ Pressure loss in metres of water = ∆pγ

=×

×114 03 1 000

9 81 1 000. ,

. ,= 1162. m

6.7.4 Sludge Piping

In sewage treatment plants, sludge piping should normally not be less than 150mm in diameter to prevent blockages. In exceptional circumstances, glass-lined pipe of smaller diameter has been used successfully. Because of theirgreater risk of blockage, gravity sludge withdrawal lines should not be lessthan 200 mm in diameter. Pump connections should not be smaller than 100mm in diameter. Instead of elbows in the line, it is good practice to installcleanouts in the form of plugged tees or crosses so that the lines can be roddedif necessary.

Velocities in the piping should be between 1.5 and 1.8 m/sec and, notwithstanding the minimum sizes, the pipe should be sized to maintain thesevelocities.

Grease has a tendency to coat the inside of piping used for primary sludge andscum. Most often, this is much more of a problem in large sewage treatmentplants than in small plants. Grease accumulation results in a decreasedeffective diameter and a consequent large increase in pumping head. Thebuildup of head occurs more slowly in systems where more dilute sludges arepumped. In some plants, specific provision is made for melting grease bycirculating hot water, steam, or digester supernatant through the main sludgelines.

Friction losses are usually relatively low in sewage treatment plants becausethe pipe lengths are relatively short. There is, accordingly, little difficulty inproviding an ample safety factor.



In long sludge lines, however, special design features should be considered.The provision of two parallel pipes should be considered, unless a single pipeshut down for maintenance for several days will not create problems.Allowance for external corrosion and pipe loads should be considered. Theprovision of facilities to supply dilution water to the lines for flushingpurposes may be necessary. Mechanical cleaning of the pipe lines may benecessary and provision should be made for the insertion of a pipe cleaner.Alternatively, or in addition, provision for steam injection for cleaning may beappropriate.

Air relief and blowoff valves at high and low points, respectively, in the pipeline may be indicated and the likelihood of water hammer phenomena,consequent to pump and/or valve operation should be considered. A discussionof water hammer is presented in Chapter 5.

6.8 Effluent Disposal Hydraulics

6.8.1 Preliminary

After treatment, sewage is either re-used or disposed of into the environment.Disposal is by far the most common and, since this is a re-entry into thehydrological cycle, it can be seen as the first step in a very indirect and long-term re-use. The most common method of disposal is by discharge anddilution into ambient waters.

It should be noted that another means of disposal is by discharge onto the landwhere the treated sewage seeps into the ground and recharges underlyinggroundwater aquifers. Part of this sewage also evaporates and, particularly indesert areas, the evaporated fraction can be substantial. Land application is notcovered in this section.

The single most important element of effluent disposal is the associatedenvironmental impact. There is an associated regulatory framework whichaffects such issues as the selection of discharge locations, the selection ofoutfall structures, and the level of treatment required. Thus, sewage treatmentand disposal are linked and cannot be considered in isolation.

This section is designed to give an introduction to effluent disposal intonatural waterways. The overall topic encompasses many areas including aterquality parameters, water quality standards and criteria, hydraulic transportprocesses such as advection and diffusion, and constituent transformationprocesses.

No attempt is made in this section to grapple with the complex numericalmathematics associated with transport and transformation processes. Manyexcellent texts are available which cover these topics. In this section, theissues of river outfalls and ocean disposal are examined qualitatively andempirically and some simple design rules are introduced.



6.8.1 River Outfalls

Many existing effluent discharges into rivers are very poorly designed. Oftenthey comprise open-ended pipes which achieve minimal initial mixing. Inshallow streams, open ended discharges on the bank may fall directly onto thewater surface, creating the potential for foaming problems.

Such problems can often be eliminated by utilising a submerged dischargepoint, farther out into the stream. Where such rivers are navigable, however,outfall design requires special attention and is likely to be closely regulated.

Rapid initial mixing of an effluent discharge into a river can be achieved witha multi-port diffuser. Such a structure discharges the effluent through a seriesof holes or ports along a pipe extending into the river. For shallow rivers, veryrapid vertical mixing is achieved over the full river depth. Turbulententrainment then draws river water into the effluent plume, promoting rapiddilution.

This situation is shown schematically in Figure 6.24, which also shows atypical elevation of a riser.

Figure 6.24: Plan and Elevation Schematic of a Typical River Diffuser

The initial dilution, S, achieved in the near field, defined as being withinapproximately one diffuser length, is given by:

S UHLQ

Q UU LHD

D D= + +� �2

1 1 22

cosα (6.47)



where U is the river velocity

H is the river depth

L is the diffuser length

UD is the discharge velocity through each port

α is the orientation of the ports above the horizontal

The diffuser length, L, is often the most important parameter as it largelydetermines the cost of the structure. Equation (6.47) is used to determine thelength of diffuser required to achieve a prescribed level of dilution. Theequation is applicable to shore-attached as well as mid-river diffusers.

The equation shows that high port discharge velocities increase dilution.However, care must be taken to ensure that there are no subsequent problemsof scour or navigation. In practice, port velocities should not exceed 3 m/sec,although this guideline may be exceeded where circumstances warrant it andespecially during infrequent high-flow events.

Figure 6.25 shows a typical river diffuser arrangement. The port spacingadopted is typically of the same order as the water depth. At the outboard endof the diffuser, a large cleanout port is provided to facilitate flushing.

Figure 6.25: Schematic of a Typical Diffuser Outfall

The primary purpose of a multi-port diffuser is to distribute the flow evenlyalong the entire length of the structure. For this reason, the discharge per portshould be as uniform as possible along the length of the diffuser. This isachieved by decreasing the diameter of the diffuser pipe in steps as shown inFigure 6.25. The detailed design to ensure an even flow distribution is based



on the so-called “manifold problem”, the details of which have been discussedby Fischer et al (1979).

The use of Equation (6.47) in practice is illustrated with Example 6.9.

Example 6.9

Determine the length and number of discharge ports for a multiport diffuserthat will provide a near-field dilution of 10 when discharging a maximum flowof 1.5 m3/sec into a river. Under low flow conditions, the river water depth is1.2 m and the current speed is 0.6 m/sec.

Note: For the shallow water conditions prevalent under low river flowconditions, the maximum discharge velocity, UD, should be lower than thevalue of 3 m/sec, recommended in the notes, to reduce the risk of bottomerosion and hazards to boaters. A value of 2 m/sec is suggested. Because ofthe shallow depth, the ports will discharge horizontally in the same directionas the river flow.

Solution

Calculate required diffuser length

SUHL

QQ U

U LHD

D D= + +� �2

1 12

2

cosα

∴ =× ××

+ +× × ×

× ×� �10

0 6 122 15

1 12 15 2 10 6 122

. ..

.. .

LL

Solve by trialL=18m

Determine required number of portsPort spacing ≈ water depth

∴ No. of ports = +1812

1.

= 16

Determine port diameter

( )QD

UD= × ×On. of portsπ 0

2

4

∴ =×

× ×� �D0

124 15

16 2.

π= 0.244m

Nearest standard size = 0.25m

Check required number of ports



NQ

UD

=× ×π 0

2

4

=×

× ×15 4

2 0 252.

.π=15.3

Select 15 ports

∴ Port velocity

=× ×

Q

ND

π2

4

=×

× ×15 4

15 0 252.

.π= 2.04m/sec (OK)

Dilution rate

S =× ×

×+ +

× × ×× ×

�

��

��

0 6 12 182 15

1 12 15 2 04 10 6 18 122

. ..

. .. .

= 10.1 (OK)

∴ Diffuser length: 18mNumber of ports: 15Port diameter: 250mmPort spacing: 1.29mPort velocity: 2.04m/secDilution (near field): 100.1

6.8.3 Ocean Disposal

Oceans and large lakes are used for effluent disposal by many communities.Provided that the outfall structure is appropriately designed, such water bodiesprovide extensive assimilation capacity.

Sewage effluent is typically carried to an offshore discharge point by a pipe ortunnel. The actual discharge may be through a single port or multi-portdiffuser.

The characteristics of the effluent plume are complicated by the densitydifference that exists between the lighter effluent and the denser sea water.The density of the effluent is dependent on its temperature and, to a lesserextent, on the suspended solids concentration.

The configuration of a typical effluent plume in the ocean is shown in Figure6.26.



Figure 6.26: Schematic of Effluent Discharge Plume in the Ocean

In the initial mixing region, also known as the “discharge near field” theeffluent is strongly buoyant and rises rapidly in the water column. This plumeentrains large amounts of ambient water, thereby diluting the effluent.Stratification of the water column means that the ambient water that is firstentrained is deep, denser water. This has the effect of reducing the plumebuoyancy as it rises into less dense ambient water. At some point during therise of the plume, its density may become equal to that of the ambient waterand the plume will rise no further.

The achievement of an equilibrium height of rise only occurs where the watercolumn is relatively strongly stratified. When the water column is notstratified, or only weakly so, the plume rises to the water surface. Beyond theinitial mixing region is the so-called “far field” where the effluent travels onambient currents and is further diluted by turbulent diffusion.

It is clear that the dilution mechanisms acting in the near field and the far fieldare very different and, for this reason, they are treated separately. The detailsof the dilution mechanisms are outside the scope of this manual and, fordetails of these and detailed design guidelines, reference should be made toRoberts et al (1989).

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Hydraulics of Sewage Treatment Plants Sec-6

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