Proceedings of the VLA Seminar

A Brief Tour of InnovativeReinforced Soil Projects:Bringing Research to Applications

Presented by

Dr. Robert LoThe University of New South WalesAustralian Defence Force Academy Campus

Cover photograph:A temporary soldier pile wall tied back by soil nails designed by Victor Li & Associates Ltd.

Published byCentre for Research & Professional DevelopmentRoom 1103 Kowloon Investment Co Ltd Building 2-12 Bute Street, Kowloon, Hong KongTel : 852-2796 1638, Fax : 852-2104 0052 Email : info@crpd-hk.com, Web Site : www.crpd-hk.com

ISBN 978-988-99028-3-4© 2009 Centre for Research & Professional DevelopmentPrinted in Hong Kong

Foreword

The VLA Seminar is organized by Victor Li & Associates Ltd (VLA). I plan to make it an annual event to promote practical application of innovative technologies in civil engineering. Speakers from Hong Kong or overseas will be invited to give presentations in the future VLA seminars.

The first VLA Seminar was held on 11 June 2009. The invited lecture was presented by Dr. Robert Lo of The University of New South Wales. It is fitting to invite my former colleague, Dr. Robert Lo, whom I truly respect as a genuine scholar and professional engineer, as the lecturer for the first VLA Seminar.

Victor LiJune 2009Hong Kong

A BRIEF TOUR OF INNOVATIVE REINFORCED SOIL PROJECTS: BRINGING RESEARCH TO APPLICATIONS

by

Dr. Robert Lo

About the Speaker: Dr. Robert Lo is Associate Professor, The University of New South Wales, Australian Defence Force Academy Campus. He graduated from The University of Hong Kong and later obtained his PhD degree from The University of New South Wales. Dr. Lo has worked in consultants in Hong Kong and Australia on a number of milestone projects before joining The University of New South Wales as an academic staff in 1986. Dr. Lo remains active in consulting work, offering his expertise in soil testing, liquefaction, reinforced soil and numerical modelling to a range of clients in Australia and Hong Kong. He has published over 140 technical papers on a wide range of topics covering soil testing, reinforced soil

and foundations. Dr. Lo maintains a research interest in geotechnical reliability and has been awarded the Thomas Telford Award for his paper on the topic of limit state design in geotechnics. He was a former member of the ISSMGE Technical Committee TC9 (reinforced soil) and currently a core member of TC39 (coastal disaster mitigation).

Proceedings of the VLA Seminar, 11 June 2009 

1

A BRIEF TOUR OF

SOME REINFORCED SOIL PROJECTS: BRINGING RESEARCH TO APPLICATIONS

S. R. Lo The University of New South Wales, ADFA Campus

Abstract: Reinforced soil system offers both proven and innovative solutions to a range of geotechnical engineering problems. While soil reinforcement technologies are well established, the drive to optimize the design solutions by making use of the range of soil reinforcement combinations continues to present engineering challenges. This paper reviews a number of projects where challenges were encountered. The application of seemingly unrelated areas of research in handling the challenges was discussed. 1. INTRODUCTION A soil structure strengthened by the introduction of structural elements into the soil mass is referred to as reinforced soil (RS), sometimes also referred to as mechanically stabilized soil (MSS). The soil reinforcement is sometimes referred to as inclusions. The reinforcement material may either be steel or geosynthetics. The former is commonly in the form of strap, mesh, and ladder. The latter takes the form of straps, sheets and grids. Reinforced soil is an effective mean in the construction of a range of fill structures such as reinforced soil wall, reinforced soil slope, and embankment on soft clay (see Figure 1a). It can also be used effectively in ensuring the stability of a steep cut by installing reinforcement bar, referred to as soil nails, progressively as the excavation progresses (Figure 1b). Soil nails are sometimes used to stabilize near-vertical cut in weak rocks. Another application is to the use of soil reinforcement to enhance the life of a pavement structure (Figure 1c). In such a case, the live load under normal working condition is dominantly cyclic. Recently soil reinforcements have been used in combination with other retaining systems such as pile wall (Figure 2a). A recent innovation proposed by Lo et al (2007) is the use of reinforcement to enhance the strength and stiffness of stone columns, which in turn function as compressive soil reinforcement in improving the load carrying capacity of soil clay (Figure 2b).

2

a) Soil reinforcement for fill structures

b) Soil reinforcement for c) Soil reinforcement for pavement steep cut slope

Figure 1: Reinforced soil applications

3

a) Soil nail and pile wall

b) Geosynthetic encased stone column

Figure 2: Combined technology in reinforced soil 2. STABILISING ROLE OF SOIL REINFORCEMENT The stabilizing role of soil reinforcement can be classified into two broad categories. In the first category, the reinforcements are placed at relatively close spacing and this produces essentially a soil-reinforcement composite as illustrated in Figure 3. If the reinforcement is essentially inextensible, the soil will be close to an at-rest state and thus cannot fail as long as the

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reinforced soil action can be maintained. However, if the soil reinforcement is extensible, the reinforcements restrict the development of lateral strain and prevent the formation of a failure mechanism. The soil around the reinforcement can be in a near-failure state although the RS system can continue to carry higher load. In the second category, the reinforcement provides a stabilizing force as illustrated in Figure 4 for a reinforced embankment on soft soil. It is also pertinent to note that both stabilising mechanisms may be in action in certain RS systems such as reinforced slope.

Figure 3: Soil-reinforcement composite

Figure 4: Provision of stabilizing force

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3. DESIGN OF REINFORCED SOIL STRUCTURE The design of reinforced soil structures generally involves determination of i) maximum reinforcement tension, Tmax, ii) resistance zone, and ii) average soil reinforcement-interface shear strength in the resistant zone. This is illustrated in Figure 5 for a RS wall, and this design paradigm is equally applicable to other forms of RS structure.

Figure 5: Design idealization

The strength of the reinforcement has to be adequate to resist the maximum reinforcement tension, Tmax, for the design load duration; and this is commonly referred to as design against reinforcement rupture. In the case of a RS wall, the connection may have a lower strength and thus an additional check based on Tc, the tensile force at connection, is also needed. The development of Tmax, requires the reinforcement being adequately anchored in the resistant zone. Therefore, the pullout resistance, Rp, must exceed Tmax by a safety margin. This is commonly referred to as design against reinforcement pullout. In the case of a RS wall, these two design checks ensure “internal stability”, i.e. the reinforcements “bind” the soil into a coherent mass that acts as a gravity retaining wall. The philosophical issue on how safety margins are to be built in, whether by a global safety factor or

6

by a series of partial factors, and how load combinations should be assessed, is not addressed by this article. Simplified calculation models for the determination of Tmax and Rp have been incorporated in a range of design codes and guidelines, for example Geoguide 6 (GEO, 2002). However, it needs to be emphasised that certain form of reinforced soil structures are highly indeterminate and the simplified equations may only be applicable for simple configurations. The very nature of reinforced soil and the availability of a wide range of reinforcements open the door for innovative design. 4. PULLOUT RESISTANCE OF STRAP REINFORCEMENT The expression for calculating pullout resistance can be expressed in a generalised form as: (1) p o pR f Aσ= where σo is the average normal stress acting on the soil-reinforcement interface, Ap is the interface area and f referred to as friction factor. All parameters are for the resistant zone. For reinforcement with constant cross section, Equation (1) reduced to:

(1a) p o pR f L pσ= where Lp is the embedded length in the resistant zone, p is the overall perimeter of reinforcement. The friction factor f is defined by Equation (1) or (1a) and is related to interface friction. f is also referred to as apparent coefficient of friction, which is not the interface coefficient of friction as measured by a sliding or direct shear test. It is also sometimes written in a rather misleading form as f = tanδp, where δp is referred to as (apparent) interface friction angle in pullout mode, which is not the same as interface friction angle, δ, that should be measured by direct shear testing. In some literatures, f is expressed as αF, where α ≤ 1 because of progressive failure. Therefore, F is referred to as a basic interface friction parameter and α is the scale reduction coefficient. In the case of steel which is essentially inextensible relative to soil deformation, α = 1 and f = F. Large scale pullout testing of geosynthetic sheets and geogrids have been conducted by a number of researchers (for example Fannin & Raju 1993; Farrag et al 1993; Bergado & Chai, 1994; Abramento & Whittle, 1995) to study

7

progressive interface failure and its effect on αF. However, similar studies on geosynthetic straps are limited. It is important to note that Equation (1) or (1a) is a two-dimensional design equation. For the case of a strap reinforcement, σo should be interpreted as an average stress at the reinforcement level and may not be the same as the normal stress acting directly on the soil-strap interface. For ribbed steel strap, most design codes or guidelines give F > tanφ at low overburden stress (σo < 120 kPa), and with F being at a maximum value of ~2 tanφ at zero reinforcement depth. The most plausible explanation for F > tanφ is constrained dilatancy of soil as illustrated in Figure 6. Due to the rough ribbed surface, soil in the vicinity of the reinforcement will be in a failure state, and shear band will form, during reinforcement pullout. The soil in this shear band, although subject to intense shearing, is constrained against dilatancy by the surrounding soil. This generates additional local normal stress acting on the reinforcement surface, which leads to a failure shear stress τf given by:

(2) ( ) δσσ

σδσσδστ tan1tantan ⎟⎟⎠

⎞⎜⎜⎝

⎛+=+==

o

yoyonf

where σn is the local normal stress acting on the reinforcement, and σy is the additional normal stress generated due to constrained dilatancy within the shear band. Setting δ = φDS for ribbed steel straps, where φDS is the friction angle of the soil measured by direct shear testing, gives:

(3) DSo

yof φ

σσ

στ tan1 ⎟⎟⎠

⎞⎜⎜⎝

⎛+=

The use of φDS is an approximation as one may argue that friction angle in simple shear is more appropriate. Since σy is not reflected by a two dimensional stress analysis, its effect has to be reflected by an increased F for ribbed steel strip. Therefore we have:

(4) DSo

y tanF φσσ

⎟⎟⎠

⎞⎜⎜⎝

⎛+= 1

Analytical models have been proposed, for example Milligan & Tei (1998), for predicting the increase in pullout resistance of soil nails due to

8

constrained dilatancy at low overburden stress. However, these models rely on the formation of a shear band during reinforcement pullout, and this cannot occur if the interface friction angle, tanφ, is less than tanφDS, which is the case for polyester straps.

Figure 6: Constrained dilatancy

4.1 Large scale pullout testing The polyester strap is shown in Figure 7. The load carrying element is the polyester yarns, whereas the polyethylene sheathing provides protection and a textured surface for enhancing interaction with the surrounding soil. Due to their high tensile load capacity, these straps are usually placed at a spacing of 600 to 1200 mm; and this simplifies construction logistics. However, the perimeter of a strap is usually small relative to its tensile capacity, and the pullout resistance at low overburden stress often controls the strap length embedded in the resistant zone. Therefore the possible presence (or absence) of this beneficial constrained dilatancy mechanism for polyester straps were investigated by large scale pullout testing.

9

It is pertinent to emphasised that pullout testing is not an element test. The various considerations in the design of a large scale pullout box have been discussed extensively in the literature (Palmeira & Milligan, 1989; Lo, 1990; Juran et al, 1991; Fannin & Raju, 1993). Pullout testing of geosynthetic sheets and grids were extensively reported in the literatures, for example Fannin & Raju (1993), Farrag et al (1993), Bergado & Chai (1994) and Abramento & Whittle (1995). In order to have a meaningful interpretation of the pullout test results, Raju et al (1998, 2000) considered the stress and displacement fields inside a pullout box and suggested that the pullout box design had to satisfy the following criteria.

The applied test pressure is essentially uniform. The effect of side wall friction has to be low. A sleeve is needed to isolate the reinforcement from the front zone

which is subject to a complicated stress field. This sleeve has to be flexible so that it will not interfere with the

applied stress. The exit at the front wall has to be specially designed so that

“jamming stress” will not develop.

Figure 7: Polyester strap

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SW-Series

tan φ

0.00 50.00 100.00Test Pressure (kPa)

0.00

0.50

1.00

1.50

Fric

tion

Fac

tor

data: grade 30 strapdata: grade 20 strapBest fit to all data

a) SW-series

PR-Series

tan φ

0.00 50.00 100.00Test Pressure (kPa)

0.00

0.50

1.00

1.50

2.00

Fric

tion

Fac

tor

data: grade 30 strapdata: grade 20Best fit

b) PR-series

Figure 8: Large scale pullout test results for polyester straps

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The pullout box had a design largely in line with that by Lo (1990). Several series of pullout testing were conducted to investigate the issue of constrained dilatancy for polyester straps (Lo, 1998). Select fill from actual construction projects were used in these studies. Two sets of results as presented in Figure 8a-b, clearly suggested that constrained dilatancy still prevailed during reinforcement pullout despite the absence of shear band. However, a model for calculating the effects constrained dilatancy, and hence F, without assuming shear band formation needs to be developed. 4.2 Constrained dilatancy modelling The M2 motorway project, a 20 km road corridor that connects the North Western suburbs of Sydney to 10 km north-west of the Sydney Harbour Bridge, provided the opportunity for the development and verification of a constrained dilatancy model for pullout resistance of strap reinforcement. This road corridor traverses forested bushland by means of bridges and RS walls. A total of thirty RS walls, with a total wall facing area of about 23,000 m2 were constructed under a design-and-construct contractual system. High tenacity polyester straps (Figure 7) were used exclusively as the reinforcing elements. Crushed sandstone excavated from the route yielded a sandy soil for used as the select fill. The constrained dilatancy model was developed based on the following assumptions:

Decomposition of strain into an elastic and a plastic component Simulating the normalised shear stress versus shear strain response

in simple shear by a simple relationship. Modelling the stress dilatancy relationship in simple shear by the

generalised form Rowe’s stress dilatancy relationship. Detailed derivation was presented in Lo (2003), but the key steps were summarised below. Decomposition of strain into elastic and plastic components leads to:

(5) pp

yey

py

eyy d

dd

dddd γγ

εεεεε ⎟⎟

⎠

⎞⎜⎜⎝

⎛+=+=

where superscripts “e” and “p” denote elastic and plastic strain components respectively. The elastic component was modelled by nonlinear elasticity with and with the Young’s modulus approximated by the Janbu Equation as expressed below (Lade, 1977):

12

(6) n

aae p

pSE ⎟⎟⎠

⎞⎜⎜⎝

⎛= 3σ

where Se and n are soil constants and pa is the standard atmospheric pressure in consistent units The family of τ-γp curves for shearing under a range of normal stress was approximated by a single L-γp relationship expressed as Equation (7) below:

(7) ( )[ ]pp

f

p

rHL ργ

γγ

−−Δ−+

= exp1

where H and rf are soil constants, L = τ/τf, and subscript “f” denotes failure state. The first term is in fact the hyperbolic function which has the capability of modelling the general shape of non-linear stress strain curves in the pre-failure state (Duncan & Chang 1970). The second term allows simulation of strain softening, with parameter Δ controlling the extent of softening and parameter ρ controlling the rate of strain softening. Prefailure dilatancy was modelled by Rowe’s Stress Dilatancy Equation (Rowe, 1962, 1972) as expressed Equation (8) below:

(8) KdEdE

out

in =

where dEin is the incremental plastic work input, dEout is the incremental plastic work output, and K is bounded by an upper and lower limit. For a simple shear deformation mode, . p

indE dτ γ= and . pout ndE dσ ε= . Since the

select fill is a dense sandy soil sheared in a plane strain mode, K = Kcv = tan2(π/4+φcv/2) and φcv is the friction angle at the critical void ratio state. Substituting and re-arranging gives:

(9) cvcv

np

y

KKdd ζσ

τ

γε

==⎟⎟⎠

⎞⎜⎜⎝

⎛−

where ζ = τ/σn is the shear to normal stress ratio. It is pertinent to note that the above equation is applicable to a prefailure state provided the soil is dilating.

13

For simple shear, dεy is identical to dεv, the volumetric strain increment. Therefore, the left hand side of the equation is in fact the dilatancy (dεv/dγ)p. Applying the above equation to the failure stress state in the soil gives the expression for predicting the dilatancy angle at failure, ψ.

(10) tantanpp

y fv SS

cv cvf f

ddd d K K

ε ζε φψγ γ

⎛ ⎞⎛ ⎞= − = − = =⎜ ⎟⎜ ⎟

⎝ ⎠ ⎝ ⎠

where “SS” denotes simple shear. Assuming φSS = φDS, the above equation is simplified to:

(10a) tantan DS

cvKφψ =

We can also relate the dilatancy at any state to that at failure by:

(11) tan tanp

y f

f cv f

dL

d Kε ζζ ζ ψ ψγ ζ ζ

⎛ ⎞ ⎛ ⎞⎛ ⎞⎛ ⎞− = − = =⎜ ⎟ ⎜ ⎟⎜ ⎟⎜ ⎟ ⎜ ⎟ ⎜ ⎟⎝ ⎠ ⎝ ⎠⎝ ⎠ ⎝ ⎠

The above key equations, after a series of algebric manipulations, lead to:

(12) ψγσσ tan. ⎟⎠⎞⎜

⎝⎛

⎟⎟⎠

⎞⎜⎜⎝

⎛= ∫

RP

C

pn

a

ooaey dL

pkpS

where ko is the in-situ earth pressure coefficient, “C” denotes the characteristic state that defines the onset of dilatancy, and “RP” denotes the state of reinforcement pullout which occurs at ζ = tanδ. Having the lower limit of integration defined by the characteristic state is consistent with Equations (5) to (10) being valid only after the onset of dilatancy. Equations (4) and (12) lead to:

(13) ( ) δψγσ

δσσ

tantan.1tan1 1

⎟⎟⎟⎟⎟

⎠

⎞

⎜⎜⎜⎜⎜

⎝

⎛

⎟⎠⎞⎜

⎝⎛

⎥⎥⎥⎥⎥

⎦

⎤

⎢⎢⎢⎢⎢

⎣

⎡

⎟⎟⎠

⎞⎜⎜⎝

⎛+=⎟⎟

⎠

⎞⎜⎜⎝

⎛+= ∫−

RP

C

pn

a

o

noe

o

y dL

p

kSF

Element testing provides the necessary input parameters for predicting (1+σy/σo), and thus F. It is pertinent to note that the variation of friction

14

angle with normal stress was also considered in the prediction using the following equation:

(14) m

a

o

DS

DS

p ⎟⎟⎠

⎞⎜⎜⎝

⎛=

σφ

φ

100,tantan

where subscript “100” denotes the value at σo = 100 kPa and m = −0.115 is an experimental parameter. It is reasonable to assume tanδ/tanδ100 = tanφDS/tanφDS,100. Therefore:

(14a) m

a

o

p ⎟⎟⎠

⎞⎜⎜⎝

⎛=

σδδ

100tantan

Friction factors, f, were measured by large scale pullout test for a range of test pressure. To enable a comparison between the predicted values of F with the measured f values, we compared the ratio F/F100 and f/f100, noting that f/f100 = F/F100 because the scale factor α was cancelled in dividing f by f100. The ratio F/F100 is given by the equation below:

(15) ⎟⎟⎠

⎞⎜⎜⎝

⎛

⎟⎟⎟⎟

⎠

⎞

⎜⎜⎜⎜

⎝

⎛

⎟⎠⎞

⎜⎝⎛

=100

100

100 tantan

tan

tanδδ

δ

δF

F

FF

The prediction is compared with test results in Figure 9. Reasonable agreement is achieved, although the data points are generally located above the predicted curve. This is somewhat expected because a number of conservative approximations were made in deriving the soil parameters. Furthermore, the comparison is based on the assumption that the magnitude of the α factor, hence progressive interface failure, is independent of the test pressure.

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0 20 40 60 80 100 120

σo (kPa)

0.6

0.8

1.0

1.2

1.4

1.6

1.8

2.0

2.2

2.4

2.6

2.8

3.0

3.2

3.4

3.6

αF

/ (α

F)10

0 o

r (F

/F10

0)pr

ed predicted Test data Trend line for test data

(subscript "100" to denote value at test pressure of 100 kPa)

Figure 9: Comparison between prediction and test results

4.3 Fine grained soil as select fill Recent road projects along the upper Blue Mountains area present challenges encountered in relation to the use of locally available materials as select fill. The sandstones in this area, when crushed, give soils with fines content up to 40% fines. This is significantly higher than that of the crushed sandstone from the Sydney Basin, which is generally used as select fill for reinforced earth wall construction in Sydney. Although the fines has low clay content, typically 5% or less, there were concerns, among others, about the pullout resistance of a ribbed steel strap reinforcement embedded in crushed sandstone excavated from the area. A study of the pullout resistance of this reinforcement-soil combination was reported in Lo et al (2004). The constrained dilatancy analysis presented in the previous section indicated that the dilatancy of the select fill has a significant influence on the friction factor. Therefore, special drained triaxial testing was conducted to investigate the stress-dilatancy behaviour of this crushed sandstone. The dilatant behaviour during all stages of shearing in a triaxial test can be seen more clearly with an R-D plot presented in Figure 10, where R = σ′1/σ′3, D

16

= 1 - dεvol/dε1, εvol = volumetric strain, and prime indicates effective stress. D, also referred to as the dilatancy factor, was computed by central difference approximation using data sets recorded at close time interval. The data points followed closely Rowe’s stress dilatancy equation (Rowe, 1962, 1972) and this indicates essentially a “sand behaviour”. The characteristic state, defined by the first occurrence of D = 1, occurs at a principal effective stress ratio of 3.10. Armed with this encouraging information, large scale pullout testing was then conducted. The pullout box with an overall dimension of 2m in length was used. This box was fitted with perforated partitions to create side and rear compartments which, for this study, stored water for wetting the soil. The partitions were lined with a non-woven geotextile in order to prevent loss of soil through the perforations and to promote wetting. A reticulation system was incorporated to ensure the compartments were always full of water. Pullout testing was conducted after completion of wetting. To verify that the pullout rate adopted was adequately slow, a high applied pullout force (at ~70% of the estimated pullout value) was maintained overnight. Essentially nil movement was observed during this hold stage. Details of the test arrangement and procedures are given in Lo et al (2004). Dummy pullout tests at zero applied pressure were also conducted. The maximum force measured in a dummy pullout test, Fdum, is a measure of the correction (or error) in the maximum pullout force. If the test condition is ideal, the pullout force recorded in a dummy pull, Fdum, should then be zero. A non-zero force can be attributed to resistance in the exit slit and apparent adhesion between the soil and the strip (as the soil is not 100% saturated). The corrected pullout force is thus the maximum measured force minus Fdum. Hence, in the calculation of friction factor, f, of pullout resistance, the corrected pullout force was used. The friction factor, f, is plotted against test pressure in Figure 11. Data points for the main test series were shown as open squares, whereas tests with prolonged soaking were shown as filled squares. Test data for the main series, tests P1 to P5a, followed a clearly defined trend, with f increasing with reduction in test pressure. However, the data points corresponding to prolonged wetting gave significantly lower friction factor. The lower friction factors achieved by these two tests were likely to be due to higher moisture content induced by prolonged wetting. The higher moisture content led to lowering of soil suction. This, in turn, led to reduction of apparent adhesion along the soil-reinforcement interface and reduction in stiffness of surrounding soil. The effect of the former is partially compensated by Fdum. The later will reduce the effect of constrained dilatancy.

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0 0.5 1 1.5 2

D= 1 - dεvol/dε1

1

2

3

4

5

6

R =

σ' 1

/σ' 3

Figure 10: Stress-dilatancy plot of crushed sandstone with high fines content 5. CONTRIBUTION OF NUMERICAL ANALYSIS Innovative wall configurations means that simplified design equations for the determination of Tmax may not be adequate. Although it is true that innovative RS walls can be studied by experimental means, especially with the use of centrifuge testing (Jaber & Mitchell 1990; Zornberg et al 1997), the time constraint imposed by a construction project makes numerical analysis a more attractive option. The successful use of numerical analyses in predicting performance of GRS walls have been reported by a number of researchers (Kapurapu & Bathurst 1995; Rowe & Ho 1997). However, a number of constraints inherent in most projects, particularly in design-and-construct projects, need to be addressed. As the fill is yet to be obtained from borrowed areas, it may only be possible to test “similar” soils. This means only simple soil models is appropriate. Since field data clearly suggest that the performance of RS walls may be significantly affected by the construction procedures, the simulation of construction sequence is considered as essential. A number of wall configurations and wall sections need to be studied numerically. There is virtually no time available for any significant modification of computer code. This means the numerical analysis must be conducted using well-tested, versatile software(s). A

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comprehensive design specification may also require that the safety margin implied by any innovative design is no less than that of conventional design. This means that the numerical analysis needs to be calibrated against representative conventional wall design, and this calibration includes the assessment of safety factors by numerical analysis. The above discussion is largely based on the author’s experience in the Eastern states of Australia and hence may or may not be equally applicable elsewhere. Despite these severe constraints, numerical modeling contributes significantly to innovative design of RS walls as attested by this paper.

20 40 60 80 100 120Test pressure (kPa)

0.5

1

1.5

2

2.5

3

f

trend for test P1 to P5a

Figure 11: Pullout resistance of crushed sandstone with high fines content 5.1 Tied back-to-back (TBB) wall In the Dutton Park to Port of Brisbane Rail Link project, about 80% of the reinforced soil walls consist of two walls aligned parallel at a distance of 6m apart (Figure 12a). The small distance between the two zones led to overlapping of the two reinforced zones. The conventional approach as suggested by FHWA design guidelines (Christopher et al, 1990; Elias & Christopher, 1996) is to design two walls as independent with overlapping of reinforcements. A possibly more “effective “approach is simply to connect the two walls with the same reinforcement (as illustrated in Figure

19

12b). Such a wall configuration is referred to as tied back-to-back (abbreviated as TBB) reinforced soil wall. However, existing design guidelines (GEO, 2002; Elias & Christopher, 1996) suggest that the reinforcement tension of a TBB wall can be considerably higher than that predicted by conventional calculation model, and that there can be difficulties in constructing a TBB wall. The heavy setback surcharge from the rail loading presented an additional challenge. The concern about higher reinforcement tension because of a TBB arrangement was first studied by a series of non-linear finite element analyses. Two types of soil reinforcement were studied: steel straps and high tenacity polyester straps. The results showed that the higher reinforcement tension was related to reinforcement stiffness; and considerably higher reinforcement tension would occur for metallic reinforcement system as reported. However, the range of stiffness of high tenacity polyester straps led only to a marginal increase in reinforcement tension. The final design adopted a TBB wall configuration that utilised high tenacity polyester straps as the reinforcing system. This reinforcement system overcame the construction difficulties by running the reinforcing straps between the two walls in a zig-zag fashion as shown in construction photo presented as Figure 13a and schematically illustrated in Figure 13b. To address the issue related to a lack of design guidelines for heavy setback surcharge from the railway loading, a set simple conservative design rules calibrated a series of non-linear finite element analyses. The final design was then verified again by finite element analysis. It is pertinent to note that both the reinforcement tension and horizontal displacement profile were not sensitive to the choice of soil models and soil parameters as discussed in detail in Lo et al (1996). The reinforcement tension of the most heavily loaded reinforcement layer under sustained loading condition was monitored at completion of wall construction, and field measurements agreed well with finite element prediction. The cost effectiveness of the TBB wall was attested by its use in a subsequent infrastructure project which involved the construction of walls up to 8.5 m in height.

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a) General arrangement

b) Schematic

Figure 12: Tied back-to-back RS Walls

21

a) Construction photo

reinf straps

attachment see detail "A"

Concrete facing panel

strap

attachment pin

attachment loop

conc

rete

pan

el

Detail "A": attachment b) Schematic reinforcement arrangement

Figure 13: Zigzag reinforcement arrangement for TBB wall

5.2 Modular block wall A 12 m high modular block wall was used to support the end span of a major bridge structure as shown in Figure 14. This wall configuration was selected on the basis of simplicity in construction, ability to accommodate differential settlement and aesthetics. The general arrangement of the wall

22

design and instrumentation was presented in Figure 15. The wall consists of four tiers. Each tier has a height in the range of 2.2 to 2.95 m, and the setback distance between tiers is ~2.0m. A bridge sill beam sits on the top tier, thus giving a total wall height of about 12m. The overall dimensions of a block are 315 mm deep by 200 mm high. HDPE geogrids were used as the soil reinforcements. The fill is a fine sand with a maximum dry density of ~1.6 t/m3. The ground conditions consist of 1 to 3 m of loose silty sand overlying 7 to 10 m of loose to medium dense silty sand. Sandstone bedrock is at approximately 13m depth. The loose sand layer contains pockets and/or lenses of soft silty clay. These silty clay pockets/lenses, although had not been located accurately, was considered to have only slight influence on the overall drainage condition of the foundation material; but may lead to some time-dependent, and possibly differential, settlement. The top 1m was however replaced with compacted sand over the front 7 m (see Figure 15).

Figure 14: Photo of modular block wall

23

ReinforcedZone

RL 1.70

SAND FILL

I-1

COMPACTED

I-2

2950

2400

2200

2250

~2200

2000 2000 2000 2500 ~2100 5000

SEGMENTALBLOCKS

SILL BEAM

GENERALFILL

2000

7000

RL 2.05

SR80@600

SR110@600

SR110@600

SR110@600

SR110@400

EPC-1 SM-3 EPC-2 SM-4

LEGEN

Re

loa

Ea

Set

HPC -3

HPC -2

HPC -1

reinforcemnet layer with load bolts and strain gaugessettlement markerearth pressure cell

Figure 15: General Arrangement and instrumentation of modular block wall The wall is classified as a stacked wall in accordance with design guidelines of AASHTO (1997) or FHWA (Christopher et al, 1990; Elias & Christopher, 1996). The essentials are reported here for the sake of completeness. The initial design was conducted using limit equilibrium analysis. In view of the unusual form of this abutment wall, FLAC (Itasca, 1998) analysis was also conducted. Details of the initial analysis were reported in Won et al (1996). These idealizations were largely dictated by the computational resources available at the time of initial design. Recognising the possible limitations of the initial analysis, the reinforcement layout adopted was conservative relative to the outcome of these analyses. The wall was instrumented with earth pressure cells, loads bolts and strain gauges along three reinforcement levels, settlement marker, horizontal profile gauges (HPG-1 to HPG-3 of Figure 15) to give near-continuous settlement profile, and inclinometers (I-1 to I-2 of Figure 15) to monitor horizontal displacements. Field measurements were taken over a long period of time, in general over 40 months. This is considered to be essential as the analysis was based on a drained condition. The monitoring results were analysed in detail by Lo (2005).

24

The measurements of the earth pressure cell at foundation level agree well with those predicted by the initial FLAC analysis. Two settlement markers, SM-3 and SM-4, were located at foundation level and to either side of the sill beam (see Figure 15). Their recorded settlements are presented in Figure 16. These two settlement markers gave essentially the same response. The maximum settlement (at 41 month) was about 90 mm and most of the settlement (~80%) was fully developed when bridge girders were constructed. The settlement profiles (from HPG data) at about 40 months after construction were presented in Figure 17. HPG-1 gives a maximum settlement of about 65 mm. Note that readings HPG-1 could be about 10 mm lower than that of settlement plate because the initial readings of HPG-1 were relative to their were taken about a month after filling began. Furthermore, HPG values taken relative to a reference point located a short distance outside the embankment footprint which may also settle slightly. The initial FLAC analysis can yield a settlement of 70 mm by assigning an extremely low Young’s modulus to the foundation materials. Horizontal displacements inferred from inclinometers I-1 and I-2 were presented in Figure 18. I-1 gave 30mm lateral movement at foundation level whereas I-2 gave even higher movement, as ~60mm at foundation level increasing to ~75 mm at tier-3. Details analysis of the observed movement patterns were presented in Lo (2005).

0 500 1000 1500time (day)

0

20

40

60

80

100

Settl

emen

t (m

m)

SM-3SM-4

bridge girders constructed

Figure 16: Development of settlement with time

25

Distance from wall face of lowest tier (m)

Set

tlem

ent

(mm

)

0 5 10 15 20 25

0

50

100

wal

l fa

ceof tier

1

wal

l fa

ceof tier

4

End o

f

rein

forc

ed z

one

HPG -3

HPG -2

HPG -1

tier

1 t

o 3

Sill

bea

m

Figure 17: Settlement profile from HPG

Figure 18: Lateral ground movements

foundation level

0 50 100horizontal displacement (mm)

25

20

15

10

5

0

Dep

th (

m)

9 month41 month

(b) I-2

0 10 20 30 40Horizontal displacement (mm)

20

15

10

5

0

Dep

th (

m)

at 10 monthsat 41 months

foundation level

(a) I-1

26

Estimate design parameters

Perform analysis to working condition

Computed movement ≥observed ?

Develop numerical model

Continue analysis to collapse

Adequate safety margin ?

OK

Refined parameters?

Additional investigation to refine parameters

RSS need strengthening

Y

N

N

Y

N

Y

Figure 19: Overall strategy

In summary, the significantly smaller movement, relative to observed values, predicted by the initial analysis presented some concern. A detailed examination of the initial analysis reveals a number of weaknesses in the initial analysis, some of which stiffened the system. In view of the significant increase in computational resources available during this study and the absence of time constraint, a series of improved analyses were conducted to investigate the status of this RS wall. The main challenge was to develop a “single” numerical model that can:

predict, conservatively, the observed high movements at working condition,

continue beyond working condition till collapse or the ultimate design condition so as to assess the safety margin in the design.

A numerical model satisfying the above two criteria, referred to as the reference model, will then be able to “feed” the observed higher movement for a working condition into the assessment of safety margin. The overall strategy was presented in Figure 19. For the context of this article, a numerical model includes location of boundaries and selecting practical aspects that need to be idealized.

27

Details of the reference model were presented in Lo (2005). In essence the numerical model that satisfies the two criteria listed above has the following attributes.

Side boundary in front of the wall foundation was placed at 35 m, a distance of ~2.5 times the thickness of the foundation soil, away from wall toe.

The construction sequence was modelled closely in a layer-by-layer manner.

The Young’s modulus of soil was modelled with the Duncan-Chang equation rather than a constant value. The use of plasticity formulation for representing Mohr Coulomb failure was still retained.

The inherent horizontal sliding planes between blocks was modelled using ubiquitous elements, which are elastic-plastic 2D elements with potential planes of weakness defined by reduced friction and cohesion values, the latter representing the equivalent contribution of dowel bars between blocks.

Technique for objectively dealing with potential numerical problem associated with large near-failure zones as discussed in Lo (2001) was implemented.

The corresponding parameters for the reference model are listed in Table 1. The predicted vertical and horizontal displacements are presented in Figures 20 and 21 respectively. The predicted displacements were significantly higher than the monitored value. The maximum predicted settlement at foundation level was 120 mm with the exception of a small zone being at the maximum value of 140mm. But the maximum measured settlement was about 80mm. The maximum horizontal deflection was 175 mm but the measured value was ~75mm. A close examination of the movements along the wall facing indicated some slipping along the joint planes. However, there is no evidence of such sliding in the field observation. Hence such predicted sliding was probably due to conservative modelling of the discrete nature of blocks by using ubiquitous elements. The predicted movements being higher than measured values is somewhat expected. The reference model inherently errs on the conservative side because of the “design intention” of the analysis: to conservatively assess the safety margin of the wall against any complicated failure mechanism but with the obserbved high movements under working condition embeded in the model.

28

Table 1: Soil Parameters for modular block wall Foundation Soil Parameter Fill

Loose Sand Medium Sand bulk density (t/m3)

φ (deg) c (kPa) ν (deg)

K n rf

1.75 33 0 5

800 1

0.9

1.8 30 0 0

400 1

0.9

1.8 33 0 0

600 1

0.9 Since the working condition was being conservatively predicted, continuation of the same analysis to a collapse condition will give conservative prediction of safety margin, hence satisfying the “design intension” in line with the limit state design principle discussed in Lo & Lee (2002). This is achieved by incrementally:

increasing the live loading to the factored value for ultimate limit

state check. reducing the strength parameter of the fill to a factored value as

prescribed in design specification R57 (2001). reducing the strength parameters of the foundation soils to either

factored values as defined in R57 (2001) or when collapse was detected, whichever occurred first.

The analysis proceeded to the ultimate (factored) condition without “ill-conditioning”. Therefore, the wall has adequate safety margin against any potential complicated mode of overall instability, and one can conclude that the numerical modelling unambiguously implied that the RS wall has adequate safety margin. Recognising that the reference analysis is inherently conservative, a parametric study was conducted to examine the effects of less conservative assumptions on strength parameters for backfill and/or less compressible ground condition. Details of such a study were reported in Lo (2005). From a practical engineering point of view, less conservative assumptions need to be justified by more detailed site investigation and laboratory testing. However, once the stability status of the wall has been established to be adequate, it becomes difficult to convince the owner of the wall to fund additional site investigation and laboratory testing.

29

Figure 20: Computed settlement at working load

Figure 21: Computed horizontal movement at working load

30

5.3 Multiple failure modes Challenges were encountered in the design of a RS wall subject to concentrated setback loads as illustrated in Figure 22. The reinforcement elements were closely spaced geosynthetic straps. This is a wall constructed for a mining operation and was subject to two concentrated live loads. The front load was distributed over 1.4m by a relieving slab, whereas the rear load was assumed to act over a narrow width of 0.4m. The behaviour of this wall was studied numerically by Lo (2003a). A series of non-linear FLAC analysis, with the initial object of assessing the reinforcement tension, was conducted. For this particular wall, the top reinforcement level was subject to heavy tension because of the concentrated loadings. Therefore, the discussion will only be confined to the top reinforcement level.

General Fill

Select Fill

relieving slab

precastwall panels

5.5

m

4.0 m 2.0 m Foundation Soil

Reinforcement (geosynthetic straps)

Concentrated Line Load

2.7 m Concentrated Patch Load

Figure 22: RS wall for mining operation

A layer-by-layer construction of the wall was modeled. The flexural stiffness of the facing was assigned a value of about 10% of the uncracked moment of inertia to take into account articulated panel construction. The axial compressibility of the facing was reduced to about 20% of the gross value to take into account the use of compressible strips between facing panels (Lo, 2001). The soil reinforcements were modelled as elastic members. The soils were modelled as Mohr-Coulomb elastic-plastic

31

material following a non-associative flow rule. Relevant geotechnical parameters (non-factored values) are listed in Table 2. These parameters were cautiously selected based on direct shear testing and construction specification. Neglecting any beneficial effects from increase in normal stress due to the rear concentrated load, the interface strength was simply modelled by the use of “bond values”, Sb, calculated by the following equation (16) p'fS vob σ=

where σ′vo was the in-situ effective overburden stress, p was the reinforcement perimeter per metre width of the wall, and Sb varied with reinforcement level. In the analyses, Sb calculated as per Equation (16) is the input property of an element. However, the interpretation of the calculated reinforcement tension at working condition for design against rupture and pullout turned out to be problematic. Therefore the analysis were brought to the factored condition in order to ensure the wall has adequate safety margin against reinforcement rupture and pullout, ie the factored design tension was assessed directly from the FLAC analysis.

Table 2: Parameters for RS wall subjected to concentrated setback load

Parameters Non-factored Unit weight (kN/m3) Friction angle (o) - Select fill - General fill Dilatancy angle (o) - Select fill - General fill

20 40 33 5 0

The design friction angles at ULS were based on a partial factor of 1.15 on tanφ, which gave values close to critical states values, and thus conforming to the rationale as discussed in Mak & Lo (1996). Unless stated to the contrary, S*b = Sb/1.25, i.e. the partial factor for Sb was 1.25. This value is similar to that recommended in R57 (2001). The design tension (for ULS) of the top reinforcement level was obtained by a separate “factored” analysis based on the following partial factors with:

Partial load factor for self-weight of soil taken as unity. Partial factors for strength parameters applied at the beginning of the

analysis.

32

Live load applied after completion of construction and then increased to the design value with a partial load factor of two.

This analysis was completed without any displacement run-off or other numerical signs of instability. It gave T* =11.1 kN/m, where T is the maximum reinforcement tension and * indicate design value obtained from a factored analysis. This appears to give the factored tensile strength for design against reinforcement rupture. However, detailed examination of the analysis results revealed significant interface yielding. Since the interface yielding could have an effect of limiting the reinforcement tension, a second factored analysis was conducted by setting S*b = Sb, i.e. partial factor on Sb was set to unity. This reduced the extent of interface yielding and gave a higher reinforcement tension of T* =12.4 kN/m. Since the friction factors, and hence Sb, tended to be conservative, the actual interface strength could be higher than the non-factored value. A third factored analysis was conducted by setting S*b = 2Sb, a not unreasonable scenario, and this suppressed all interface yielding. The resultant T* value is 14.5 kN/m, nearly 35% higher than that calculated in the first factored analysis. Therefore, “mechanistic factoring down” of all strength parameters in a complicated soil structure interaction system may lead to under-estimation of design reinforcement tension. The factored reinforcement tension, T*, is needed for design against two failure modes: reinforcement rupture and reinforcement pullout. The critical “design scenario” for reinforcement rupture is different from that of reinforcement pullout. The first scenario is given by an analysis that suppresses interface yielding. The most appropriate design scenario for reinforcement pullout is the first analysis that applies the partial factor of 1.25 on f. Since the analysis can be completed satisfactorily (with factored geotechnical and interface parameters) without any runoff in wall displacement or numerical signs of instability, it was geotechnically stable.

6. REINFORCED EMBANKMENT ON SOFT SOIL A section of a freeway extension is a 300 m long embankment constructed a swamp. The embankment has a maximum elevation of 5.5m and a base width of ~60 m. The embankment cross-section is shown in Figure 23. The foundation clay has a high water content in the range of 77% to 99% and a very high compression index, Cc of 0.85. The settlement registered to date is ~1.8m. A number of measures were incorporated in the design of this embankment in order to enhance stability and reduce the post construction ground movements to an acceptable limit. These measures included the use

33

of wide stabilizing berm, lightweight fill (bottom ash), prefabricated vertical drains, surcharging, and staged construction and geogrid reinforcement at the base of the embankment. The embankment was extensively instrumented with settlement profilers (HPG), piezometers, inclinometers, earth pressure cells and reinforcement force/strain sensors (Figure 23). Pressure, force and strain (extensometer) sensors were monitored for about 400 days, whereas the movement profilers were monitored for over 9 years. A detailed analysis on the long term performance of this embankment is presented in Lo et al (2008).

0 20 40 60 80Distance from western toe(m)

-15

-10

-5

0

5

Ele

vatio

n (m

R.L

.)

Existing Embankment

Berm(Ordinary Fill)

Embankment Core (Light weight Fill)

R.L. 5.5mR.L. 4.8m

R.L. 3.1m

R.L. 1.1m

I2.2, I2.6

I2.1

Ordinary Fill

P2.1(P1.1)

P2.2(P1.2)

P2.3(P1.3)

P2.4(P1.4)

P2.5(P1.5)

P2.6(P1.6)

P2.7(P1.7)

P2.8(P1.8)

P2.10(P1.10)

P2.11(P1.11)

P2.12(P1.12)

P2.9(P1.9)

P2.13

P2.14

West

HPG Piezometer Earth Pressure Cell Inclinometer( ) Line-1 instrumentation Geogrid with load bolts

treated by PVD

Existing Lenghan Drive

LB8 LB9 LB10 LB11 LB12

LB13 LB14

Figure 23: General Arrangement and Instrumentation of reinforced embankment

The basal reinforcement is a uniaxial flexible geogrid. The load carrying elements are longitudinal straps formed from high strength polyester yarns encapsulated in low density polyethylene. The transverse members are there to form a grid for easy installation. This geogrid has a short term strength of 200 kN/m width and a short term stiffness of 2800 kN/m width. Despite the inherent uncertainty in the monitored geogrid force, its value was significantly lower than the value inferred by stability calculation. Although both settlement and lateral movements increased significantly over the first 5 years, the reinforcement tension was essentially stationary at completion of construction. The location of maximum reinforcement tension shifted towards the centre zone near the end of construction, which is an indication

34

of gain in stability; and during this shift, the magnitude of maximum reinforcement tension decreased. This removes any concern about the long term strength of the geogrid reinforcement. One might question the role of basal reinforcement. An examination of the inclinometer readings indicated that the lateral movement reduced significantly near ground level (Figure 24). This is because of restraint from the geogrid reinforcement (Lo & Li, 1993). It is pertinent to note that the embankment fill is not an elastic material and its strength and stiffness depends on confinement or loss of confinement. This implies that the basal reinforcement may mitigate against loss of confinement and thus ensure the embankment fill will not suffer from a spreading/slumping failure.

-150 -100 -50 0

Horizontal deflection (mm)

-20

-15

-10

-5

0

Red

uce

Leve

l (m

)

42 days (3.1m)152 days (3.1m)303 days (4.2m)371 days (5.5m)547 days (5.5m)

(.) indicates approximate R.L. of embankment elevetion

natural ground level

Figure 24: Lateral movement at toe of reinforced embankment

35

7. SUMMARY A brief tour of several reinforced soil projects was presented in this paper. Although reinforced soil construction is a relatively mature technology, the opportunities for innovation are still plentiful. This is partly due to the multi-component nature of reinforced soil. A number of research tools, ranging from shear-dilatancy analysis, large scale laboratory testing, field instrumentation, and non-linear numerical analysis were applied. However, the research tool used is driven largely by the project challenge following the paradigm of bringing research to applications. ACKOWLEDGEMENT Many engineers contributed to the projects addressed in this paper. The opinions expressed in this paper are, however, solely those of the author.

36

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