Steel Construction Introduction to Design

Lecture 1B.1: Process of Design

OBJECTIVE/SCOPE

To introduce the challenge of creative design and to explain approaches by which it may be achieved.

PREREQUISITES

A general knowledge of basic applied mechanics is assumed and prior encouragement should be given to

read J E Gordon's three books [1,2,3].

RELATED LECTURES

Since this lecture deals with the process of design in general terms almost all other lectures are related to

it in some way. Those sections which are most closely associated with it are 1B:Introduction to Design,

14: Structural Systems: Buildings, 15A: Structural Systems: Offshore, 15B: Structural Systems: Bridges,

and 15C: Structural Systems: Miscellaneous

SUMMARY

The lecture begins by considering a definition of design and some objectives. It discusses how a designer

can approach a new problem in general and how a structural designer can develop a structural system. It

concludes by considering differences of emphasis in design approach for different classes of structure.

1. DESIGN OBJECTIVES

The results of successful design in structural engineering can be seen and used by everyone, see Figure 1.

The question is: how can professional designers be developed and eventually produce better designs than those previously encountered, to benefit and enhance the performance of human activities? In particular how can steel be utilised effectively in structures for:

• travelling more easily over awkward terrain, requiring bridges. • enabling basic industrial processes to function requiring, for example, machinery supports, docks

and oil rig installations. • aiding communications, requiring masts. • enclosing space within buildings, as in Figure 2.

Design is 'the process of defining the means of manufacturing a product to satisfy a required need': from the first conceptual ideas, through study of human intentions, to the detailed technical and manufacture stages, with the ideas and studies communicated with drawings, words and models.'Designers'? All people are capable of creative conceptual ideas - they are continuously processing information and making conscious imaginative choices, e.g. of the clothes they wear, of the activities they engage in, and the development of ideas they pursue, causing changes.

In structural design, prime objectives are to ensure the best possible:

• unhindered functioning of the designed artefact over a desired life-span. • safe construction system, completed on time and to the original budget cost. • imaginative and delightful solution for both users and casual observers.

These points could possibly be satisfied by either:

• simply making an exact copy of a previous artefact, or, • 're-inventing the wheel', by designing every system and component afresh.

Both these extreme approaches are unlikely to be entirely satisfactory. In the former case, the problem may well be slightly different, e.g. the previous bridge may have stimulated more traffic flow than predicted, or vehicle weights may have increased. Economic and material conditions may have changed, e.g. the cost of labour to fabricate small built-up steel elements and joints has increased compared to the production cost of large rolled or continuously welded elements; also, corrosion resistant steels have reduced maintenance costs relative to mild steel. Deficiencies of performance may have been discovered with time, e.g. vibrations may have caused fatigue failures around joints. Energy consumption conditions may have changed, e.g. relating to the global discharge of certain chemicals, the cost of production of certain materials, or the need for greater thermal control of an enclosed space. Finally, too much repetition of a visual solution may have induced boredom and adverse cultural response, e.g. every adjacent building is produced in the "Post Modern Style".

With the latter approach, 'life is often just too short' to achieve the optimal solution whilst the client frets.... Civil and structural engineering projects are usually large and occur infrequently, so a disenchanted client will not make a second invitation. Realisation of new theoretical ideas and innovations invariably takes much time; history shows this repeatedly. Thus methodical analysis of potential risks and errors must temper the pioneering enthusiast's flair.

Positive creative solutions must be achieved for all aspects of every new problem. The solutions will incorporate components from the extremes above, both of fundamental principles and recent developments. However, throughout the Design Process it is prudent to maintain a clear grasp of final objectives and utilise relatively simple technical means and solutions.

2. HOW DOES THE DESIGNER APPROACH HIS NEW TASK?

At the outset of a new task an "instant of blind panic" may occur. There are a variety of Design Methods to help progress [4, 5] with the new task, but the following methodical approach is suggested:

1. Recognise that a challenge exists and clearly define the overall objectives for a design, see Figure 3.

2. Research around the task and investigate likely relevant information (Analysis). 3. Evolve possible solutions to the task (Synthesis). 4. Decide on, and refine, the best solution (Evaluation), establishing clear priorities for action (in

terms of manufacture, construction, operation and maintenance). 5. Communicate decisions to others involved in the task.

At the outset, these five phases appear as a simple linear chain; in fact the design process is highly complex, as all factors in the design are interdependent to a greater or lesser degree. Hence there will be many steps and loops within and between the phases, as seen in Figure 4. The first rapid passage through phases 1, 2 and 3 will decide if there is 'any problem', e.g. is the likely traffic flow adequate to justify a convenient but high cost bridge?

All factors and combinations must be explored comprehensively from idea to detail, with many compromises having to be finely balanced to achieve a feasible solution. Ideas may be developed: verbally, e.g 'brainstorming' or Edward de Bono's 'lateral thinking' approaches [6], graphically, numerically or physically. Always qualitative assessment should proceed quantitative evaluation.

The starting point for Analysis may thus be the designer's current preconceived notion or visual imagination, but the Synthesis will reveal the flexibility of his mind to assimilate new ideas critically, free of preconception.

A designer can prepare himself for the compromises and inversions of thought and interaction with other members of the Design Team leading to successful synthesis, through 'Roleplay Games', e.g. see 'The Monkey House' game, in Appendix 1.

3. HOW DOES THE DESIGNER DEVELOP HIS STRUCTURAL SYSTEM?

An example of structural design, and the various decision phases, will be briefly considered for a simple two-lorry garage building with an office, toilet and tea room, shown completed in Figure 2. It is assumed in this hypothetical case that an initial decision has already been made by the client to have this set of requirements designed and built.

3.1 Pose an Initial Concept that may well Satisfy the Functions

It is invariably the best idea to start by looking at the functions (performance) required and their relationships. Make a list of individual functions; then generate a 'bubble' (or flow) diagram of relationships between different functional areas to decide possible interconnections and locations, see Figure 5. Find, or assume, suitable plan areas and minimum clear heights of each three-dimensional 'volume of space'. A possible plan layout may then be indicated, noting any particular complications of the site, e.g. plan shape, proximity of old buildings, slope or soil consistency.

Many other plan arrangements will be possible and should be considered quickly at this phase.

The requirements of each 'volume of space' and its interfaces must be examined for all functional, cost and aesthetic criteria, e.g. what structural applied live loads must be resisted; what heating, ventilating, lighting and acoustic requirements are likely to be desired, see Figure 6.

The main criteria can easily be recognised and then followed up and tested by numerical assessment. Incompatibilities may be 'designed out' by re-arranging the planned spaces or making other compromises, see Figure 7, e.g. would you accept an office telephone being very close to the workshop drill or lorry engine, without any acoustic insulation?

Prepare a set of initial assumptions for possible materials and the structural 'Frame', 'Planar' or 'Membrane' load-bearing system [7] that might be compatible with the 'volumes of space' as shown in Figure 8. These assumptions will be based on previous knowledge and understanding of actual constructions[8-13] or structural theory, see Figure 9 a, b, as well as the current availability of materials and skills. Initial consultations may be needed with suppliers and fabricators, e.g. for large quantities or special qualities of steel.

Steelwork, with its properties of strength, isotropy and stiffness, and its straight and compact linear elements, lends itself to 'Frame' systems, see Figure 9 c-e, which gather and transfer the major structural loads as directly as possible to the foundations, as a tree gathers loads from its leaves through branches and main trunk to the roots.

Next (and continuously) elucidate and test your ideas by making quick 3D sketches, or simple physical models, to explore the likely compatibility and aesthetic impact.

A range of stimulating evocative patterns viewed at different distances from, all around, and inside the buildings must be developed:

Long range the skyline silhouette or "landscape" pattern

Middle distance when the whole built object can be seen

Close up when a detail is clearly seen

Very close when the texture of the materials can be seen.

All these conditions should be satisfied, and especially for very large buildings for most of the time. Deficiencies may be made up in some people's minds if their social conditions change for the better or natural or changing phenomena occur, e.g. the rays of the setting sun suddenly give a completely different colour appearance or after sunset the interior lighting creates patterns previously unnoticed.

Form, colour, warmth and definition can be achieved with skilful use of steel, especially with "human scale" elements though repetition will soon induce boredom; but only as part of the complete sensory experience which must include elegant solutions to all aspects - especially those easily visible - of the total building design.

It is very important that all principal specialists (architects, engineers for structure and environmental services, and also major suppliers and contractors who should all have common education and understanding of basic design principles) collaborate and communicate freely with each other - also with the client - at this conceptual design phase. Bad initial decisions cannot subsequently be easily and cheaply rectified at the more detailed design phases.

Be prepared to modify the concept readily (use 4B pencils) and work quickly. Timescale for an initial structural design concept: seconds/minutes. But hours will be needed for discussion and communication with others in researching an initial complete design idea.

3.2 Recognise the Main Structural Systems and Contemplate the Necessary Strength and Stiffness

Consider the applied live loads from roofs, floors or walls, and trace the 'load paths' through the integral 3D array of elements to the foundations, see Figure 10.

If the roof is assumed to be profiled steel decking, the rainwater should run to the sides, and a manufacturers' data table will indicate both the slope angle to be provided (4° - 6° minimum) and the

secondary beam (purlin) spacing required, e.g. commonly 1,4m - 2,6m. The purlins must be supported, e.g. commonly 3m - 8m, by a sloped main beam or truss, usually spanning the shorter direction in plan, and supported by columns stabilised in three dimensions.

Wind loads on the longer side of the building can be resisted by cladding that spans directly to the main columns, or onto sidewall rails spanning between columns. The columns could resist overturning by:

• cross-bracing (in this case the large entry door would be impeded). • or rigidly fixing the columns to the foundation bases ("linked cantilevers"); can the soil resist the

extra overturning effect at the base? • or rigidly fixing the tops of the columns to the main beams (creating 'portals') and giving smaller,

cheaper "pin" base foundations.

Wind loads on the open short side of the building can be resisted by the opening door spanning top or bottom, or side to side. At the closed short side the wind loads can be resisted by cladding that either spans directly between secondary end wall columns, or onto rails to these columns.

At both ends of the building, longitudinal forces are likely to be induced at the tops of the columns. Trussed bracing can be introduced, usually at both ends of the roof plate, to transfer these loads to the tops of a column bay on the long side - which must then be braced to the ground.

Identify the prime force actions (compression C; tension T; bending B) in the elements and the likely forms of overall and element deflections for all applied loadings both separately and when combined.

It is always useful to have the elements drawn to an approximate scale, which can be done using manufacturers' data tables for decking and cladding, from observations of existing similar buildings, or using 'Rules of Thumb', e.g. the span/depth ratio for a simply-supported beam equals about 20 for uniform light roof loading, see Figure 11.

At this phase the structural design becomes more definite (use B pencil) and takes longer. Timescale: minutes.

3.3 Assess Loads Accurately and Estimate Sizes of Main Elements

Establish the dead load of the construction and, with the live loads, calculate the following, see Figure 12:

• beam reactions and column loads (taking half the span to either side of an internal column). • maximum bending moments, e.g. wL2/8 for a simply supported beam, under uniform load. • maximum shearing forces in beams. • deflection values, e.g. 5/384 wL4/EI for a simply supported beam with uniform load.

The size of columns carrying little moment can be estimated from Safe Load Tables by using a suitable effective length. Significant bending moments should be allowed for by a suitable increase, i.e. twice or more, in section modulus for the axis of bending.

Beam sizes should be estimated by checking bending strength and stiffness under limiting deflections. Structure/service duct or pipe integration may require beams to be as shallow as possible, or deeper and with holes in the web.

Likely jointing methods must be considered carefully: is the beam to be simply supported or fully continuous and what are the fabrication, erection and cost implications?

Structural calculations are now being performed (use HB pencil with slide rule, simple calculator or computer) and the time involved is more significant. Timescale: minutes/hours.

3.4 Full Structural Analysis, using Estimated Element Sizes with Suitable Modelling of Joints, Related to Actual Details

Carry out a full structural analysis of the framework, either elastically or plastically. A computer may well be used, though some established 'hand' techniques will often prove adequate; the former is appropriate when accurate deflections are required, see Figure 13.

For the analysis of statically indeterminate structures, an initial estimate of element stiffnesses (I) and joint rigidity must be determined by the third phase above, before it is possible to find the disposition of bending moments and deflections. If subsequent checking of the design of elements leads to significant changes in element stiffness, the analysis will have to be repeated. The role of the individual element flanges and web in resiting local forces within connections must also be considered very carefully when determining final element sizes. Excessive stiffening to light sections can be prohibitively expensive.

The analysis cannot be completed without careful structural integration and consideration of the compatibility of the entire construction system including its fabrication details.

Element joints will usually be prepared in the factory using welding, with bolts usually completing joints of large untransportable elements at site. Bracings, deckings and claddings will usually be fixed on site with bolts or self-tapping screws. It is important to remember that failures most frequently arise from poor jointing, details and their integration.

The structural calculations and details are now progressing (use HB pencil with slide rule, calculators and computers). Timescale: hours/days.

Iteration of phases 1-4 above will undoubtedly be required, in particular to ensure that the early structural decisions are compatible with the subsequent investigations concerning the functional, environment, cost and aesthetic aspects. The effect of any change must be considered throughout the complete design. Changes usually necessitate a partial 're-design'.

3.5 Communicate Design Intentions through Drawings and Specifications

Prepare detail drawings and specifications for contractors' tenders, see Figure 14. Iteration of the design may again be necessary, due to variations in contractors' prices and/or preferred methods, e.g. welding equipment available, difficulties in handling steelwork in the fabricating shop or for transportation and erection. Changes and innovations in the design must be communicated and specified very carefully and explicitly.

In many cases it is common practice for a Consulting Structural Engineer to prepare preliminary designs with choice of main sections, leaving a Steelwork Fabricator to complete the detailed design and jointing system, before checking by the Consultant.

The structural design is now being finalised (use 2 to 4 H pencils and pens, or computers). Timescale: days/weeks.

3.6 Supervise the Execution Operation

Stability of the structure must be ensured at all stages of the execution, see Figure 15. High quality components and skilled erectors must be available at the right place and time, calling for very careful organisation. If 'all goes to plan' every piece will fit into the complete jigsaw.

The design ideas are now being put into operation (use gumboots). Timescale: weeks/months.

3.7 Conduct Regular Maintenance

Only regular maintenance already thoroughly planned into the design will be needed, with occasional change and renovation needed with change of use or occupation. Correction of design faults due to innovation and errors should not be needed.

This is the operation phase. (Use a serene outlook on life!) Timescale: years/decades.

3.8 Differences of Emphasis in Design Approach Compared to that of a Medium Sized Building

3.8.1 Single houses

Most "traditionally" built timber and masonry houses include some standard steel elements, e.g. hot-rolled steel beams to span larger rooms and support walls, hollow section columns for stair flights, cold-rolled lintels over window openings, stainless steel wall ties and straps, also nails, screws and truss-rafter nail plates.

Cold-rolled galvanised or stainless steel sections can be made up into truss-rafters and replace timber in repetitive conditions. Similar sections can be made up as stud walls, but fire protection of the thin-walled sections will require careful attention, especially for multi-storey houses.

A main steel structural frame may be used for houses, but integration of services, thermal control, fire protection in multi-storeys, corrosion and fabrication costs of elegant jointing must be designed appropriately. Various types of profiled or composite panel cladding can be used for the exterior.

3.8.2 Bridges

The magnitudes of gravity loading are often relatively greater in bridges, and particular load patterns need to be assessed; also trains of moving wheel loads will occur giving marked dynamic effects. Dynamic

effects of wind loading are significant in long-span structures. Accessibility of site, constructability of massive foundations, type of deck structure and regular maintenance cost will govern the system adopted. Aesthetics for users and other observers are important; long distance scale should be appropriately slender but psychologically strong; careful attention is needed for fairly close viewing of abutments and deck underside.

3.8.3 Offshore oil rigs

The scale of the whole operation will be very many times that of an onshore building. Gravity loading, wind speeds, wave heights and depth of water are significant design parameters for structure size and stability (here larger elements cause larger wind and wave loads). The scale of the structure also poses special problems for fabrication control, floating out, anchorage at depth by divers and, not least, cost, see Figure 1. Later when the design life is complete, the problems of dismantling should be easy, if considered during the initial design.

4. CONCLUDING SUMMARY

• This lecture introduces the challenge of creative design and suggests a holistic strategy for designing structural steelwork. It seeks to answer questions about what a designer is trying the achieve and how he can start putting pen to paper. It illustrates how a successful design is iterated, through qualitative ideas to quantitative verification and finally execution.

• Creative and imaginative design of structures is most challenging and fun - now try it and gain confidence for yourself. Do not be afraid of making mistakes. They will only be eliminated by repeating and exploring many other solutions. Make sure the design is right before it is built, using your own personal in-built checking mechanisms.

5. REFERENCES

[1] Gordon, J. E. 'The New Science of Strong Materials', Pelican. [2] Gordon, J. E. 'Structures', Pelican. [3] Gordon, J. E. 'The Science of Structures and Materials', Scientific American Library, 1988. [4] Jones, J. C. 'Design Methods', Wiley 2nd Edition 1981. A good overview of general design methods and techniques. [5] Broadbent, G. H. 'Design in Architecture', Wiley, 1973. Chapters 2, 13, 19 and 20 useful for designing buildings. [6] De Bono, E. eg: 'Lateral Thinking' or 'Practical Thinking' or 'The Use of Lateral Thinking', Pelican. [7] LeGood, J. P, 'Principles of Structural Steelwork for Architectural Students', SCI, 1983 (Amended 1990). A general introduction and reference booklet to buildings for students. [8] Francis, A. J, 'Introducing Structures, Pergamon, 1980. A good overview text, especially Chapter 11 on Structural Design. [9] Lin, T. Y. and Stotesbury, S. D, 'Structural Concepts and Systems for Architects and Engineers', Wiley, 1981. Chapters 1-4 give a very simple and thoughtful approach to total overall structural design, especially for tallish buildings. [10] Schodek, D. L, 'Structures', Prentice Hall, 1980. Good clear introductory approach to structural understanding of simple concepts, also especially chapter 13 on structural grids and patterns for buildings. [11] Otto, F, 'Nets in Nature and Technics', Institute of Light Weight Structures, University of Stuttgart, 1975.

Just one of Otto's excellent booklets which observe patterns in nature and make or suggest possible designed forms. [12] Torroja, E, 'Philosophy of Structures', University of California Press, 1962. Still a unique source book. [13] Mainstone, R. J, 'Developments in Structural Form', Allen Lane, 1975. Excellent scholarly historical work, also chapter 16 on 'Structural Understanding and Design'. APPENDIX 1

'The Monkey House' roleplay game for a group of students at a seminar, Figure 16

Between 10 and 12 acting roles are created, one for each student in the group, to consider design requirements and interactions. Each actor sees an outline sketch plan of a possible building and has about

3 minutes to prepare his role's requirements, likes and dislikes. These requirements are propounded for about 2/3 minutes to his uninterrupting fellow participants, who note points of agreement/disagreement. When all actors have spoken, the many conflicts are then generally discussed and explored by the actors for about 30 minutes. Then the chairperson seeks a conclusion - who is The Monkey House really for?

Lecture 1B.2.1: Design Philosophies

OBJECTIVE/SCOPE:

To explain the objectives of structural design and the uncertainties which affect it; to outline how different priorities might influence the design, and to describe different approaches to quantifying the design process.

RELATED LECTURES:


Lecture 1B.3: Background to Loadings

Lecture 1B.8: Learning from Failures

Lecture 2.4: Steel Grades and Qualities

Lecture 2.5: Selection of Steel Quality

SUMMARY:

The fundamental objectives of structural design are discussed. The uncertainties associated with designing structures in terms of loading and material properties are considered. The development of structural design methods for strength and resistance is reviewed briefly and the importance of achieving structural stability is explained. Other design considerations such as deflections, vibration, force resistance and fatigue are discussed. Matters of construction and maintenance are included. The importance of considering these aspects and others, such as accommodating services and cladding costs, in developing an efficient design is emphasised. The responsibilities of the designer and the need for effective communication are considered.

1. INTRODUCTION

The precise objectives of structural design vary from one project to another. In all cases, the avoidance of collapse is an important - if not the most important - requirement and an adequate factor of safety must be provided. In this context, the structure must be designed in order to fulfil both strength and stability requirements. These concepts are illustrated in Figure 1 in which a long thin rod is subject to tension (Figure 1a) and compression (Figure 1b). In the case of tension, the load resistance of the rod is governed by strength, that is the ability of the material to carry load without rupturing. The rod can only carry this load in compression if it remains stable, i.e. it does not deform significantly in a direction perpendicular to the line of action of the applied load. The stiffness of the structure is yet another important characteristic, concerned with resistance to deformation rather than collapse. This is particulary important in the case of beams whose deflection under a particular load is related to their stiffness (Figure 1c). Large deformations are not necessarily associated with collapse, and some brittle materials, such as glass, may rupture with little prior deformation. Other considerations may also need to be included in the design process. They include: quantifiable behaviour such as deformation, fatigue, fire resistance and dynamic behaviour; considerations such as corrosion and service accommodation which may influence both detail and overall concept, but in a more qualitative way; and appearance, which is largely a subjective judgement. In addition considerations of economy are likely to be a significant influence on the great majority of

structural designs. In this context questions of speed and ease of construction, maintenance and running costs, as well as basic building costs, are all relevant. The relative importance of each of these aspects will vary depending on circumstances.

The approach to structural design is dealt with in Lecture 1B.1, which describes how the designer might begin to accommodate so many different requirements, many of which will exert conflicting pressures. In this lecture the focus is on how a satisfactory structural design can be achieved through a rational analysis of various aspects of the structure's performance. It is worth emphasising that the process of structural design can be considered as two groups of highly interrelated stages. The first group is concerned with defining the overall structural form - the type of structure, e.g. rigid frame or load bearing walls, the arrangement of structural elements (typically in terms of a structural grid), and the type of structural elements and material to be used, e.g. steel beams, columns and composite floor slabs. A high degree of creativity is required. The synthesis of a solution is developed on the basis of a broad understanding of a wide range of topics. The topics include structural and material behaviour, as well as a feel for the detailed implications of design decisions made at this stage - for instance recognising how deep a beam may need to be for a particular purpose. Formalised procedures are of little use at this stage. A satisfactory solution depends more on the creative ability of the designer.

The later stages are concerned with the more detailed sizing of structural components and the connections between them. By now the problem has become clearly defined and the process can become more

formalised. In the case of steelwork the process generally involves selecting an appropriate standard section size, although in some circumstances the designer may wish to use a non-standard cross-section which, for execution, would then need to be made up, typically by welding plates or standard sections together into plate girders or trusses.

Design regulations are largely concerned with this stage of detailed element design. Their intention is to help ensure that buildings are designed and constructed to be safe and fit for purpose. Such design legislation can vary considerably in approach. It may be based simply on performance specification, giving the designer great flexibility as to how a satisfactory solution is achieved. An early example of this is the building laws published by King Hummarabi of Babylon in about 2200BC. They are preserved as a cuneiform inscription on a clay tablet and include such provisions as 'If a builder builds a house for a man and does not make its construction firm and if the house which he has built collapses and causes the death of the owner of the house, then that builder shall be put to death. If it causes the death of the son of the owner of the house, then a son of the builder shall be put to death. If it causes the death of a slave of the owner of the house, then the builder shall give the owner a slave of equal value'. The danger, and at the same time the attraction, of such an approach is that it depends heavily on the ability of the designer. Formal constraints, based on current wisdom, are not included and the engineer has the freedom to justify the design in any way.

The other extreme is a highly prescriptive set of design rules providing 'recipes' for satisfactory solutions. Since these can incorporate the results of previous experience gained over many years, supplemented by more recent research work they might appear to be more secure. However, such an approach cannot be applied to the conceptual stages of design and there are many cases where actual circumstances faced by the designer differ somewhat from those envisaged in the rules. There is also a psychological danger that such design rules assume an 'absolute' validity and a blind faith in the results of using the rules may be adopted.

Clearly there is a role for both the above approaches. Perhaps the best approach would be achieved by specifying satisfactory performance criteria to minimise the possibility of collapse or any other type of 'failure'. Engineers should then be given the freedom to achieve the criteria in a variety of ways, but also be provided with the benefit of available data to be used if appropriate. Perhaps the most important aspect is the attitude of the engineer which should be based on simple 'common sense' and include a healthy element of scepticism of the design rules themselves.

2. UNCERTAINTIES IN STRUCTURAL DESIGN

Simply quantifying the design process, using sophisticated analytical techniques and employing powerful computers does not eliminate the uncertainties associated with structural design, although it may reduce some of them.

These uncertainties include the following:

• loading. • constitutive laws of the material. • structural modelling. • structural imperfections.

Loading is discussed in more detail in Lecture 1B.3. Although it is possible to quantify loads on a structure, it is important to recognise that in most cases these represent little more than an estimate of the likely maximum load intensity to which a structure will be exposed. Some loads, such as the self weight

of the structure, may appear to be more easily defined than others, such as wind loads or gravity waves on offshore structures. However, there is a significant degree of uncertainty associated with all loads and this should always be recognised.

Constitutive laws are typically based on the results of tests carried out on small specimens. For convenience, the mathematical representation of the behaviour, for instance in the form of a stress-strain curve, is considered in a simplified form for the purpose of structural design. In the case of steel the normal representation is linear elastic behaviour up to the yield point with plastic behaviour at higher strains (Figure 2). Although this representation provides a reasonable measure of the performance of the material, it is clearly not absolutely precise. Furthermore, any material will show a natural variability - two different samples taken from the same batch will typically fail at different stresses when tested. Compared with other materials, steel is remarkably consistent in this respect, but nevertheless variations exist and represent a further source of uncertainty.

Methods of analysing structural behaviour have advanced significantly in recent years, particularly as a result of developments in computing. Despite this, structural analysis is always based on some idealisation of the real behaviour. In some cases, such as isolated beams supported on simple bearings, the idealisation may be quite accurate. In other circumstances, however, the difference between the model and the real structure may be quite significant. One example of this is the truss which is typically assumed to have pinned joints, although the joints may in fact be quite rigid and some members may be continuous. The assumption that loadings are applied only at joint positions may be unrealistic. Whilst these simplifications may be adequate in modelling overall performance the implications, at least with regard to secondary effects, must be recognised.

Yet another source of uncertainty results from structural imperfections which are of two types: geometrical, i.e. out of straightness or lack of fit, and mechanical, i.e. residual stresses due to fabrication procedures or inhomogenities in the material properties. It is not possible to manufacture steel sections to

absolute dimensions - wear on machinery and inevitable variations in the manufacturing process will lead to small variations which must be recognised. In the same way, although steel construction is carried out to much tighter tolerances than for most other structural materials, some variations (for instance in the alignment of individual members) will occur (Figure 3).

In adopting a quantified approach to structural design, all these uncertainties must be recognised, and taken into account. They are allowed for by the following means:

• specifying load levels which, based on previous experience, represent the worst conditions which might relate to a particular structural type.

• specifying a sampling procedure, a test plan and limits on material properties. • specifying limits or tolerances for both manufacture and execution. • using appropriate methods of analysis, whilst recognising the difference between real and

idealised behaviour.

These measures do not eliminate the uncertainties but simply help to control them within defined bounds.

3. DESIGNING TO AVOID COLLAPSE

3.1 Historical Background

Structural design is not something which is new. Ever since man started building - dwellings, places of worship, bridges - some design philosophy has been followed, albeit often unconsciously. For many centuries the basis of design was simply to copy previous "designs". Where "new developments" or modifications were introduced, trial and error techniques were all that was available. As a result many

structures were built, or partially built only to collapse or perform inadequately. Yet these failures did have a positive value in that they contributed to the fund of knowledge about what is workable and what is not.

This unscientific approach persisted for many centuries. Indeed it still forms part of the design approach adopted today. Rules of thumb and empirical design recommendations are frequently used, and these are largely based on previous experience. Nor is structural engineering today totally free of failures, despite the apparent sophistication of design methods and the power of computers. The dramatic box girder bridge collapses in the early 1970s were a grim reminder of what can happen if new developments are too far ahead of existing experience.

The emergence of new materials, notably cast and wrought iron, required a new approach and the development of more scientific methods. The new approach included testing, both of samples of the material and proof testing of structural components and assemblies. New concepts too were sometimes justified in this way, for instance in the case of the Forth Rail Bridge.

The first moves to rationalise structural design in a quantitative way came at the beginning of the 19th century with the development of elastic analysis. This type of analysis allowed engineers to determine the effect (on individual structural components) of forces applied to a complete structure.

Testing of materials provided information concerning strength and, in the case of iron and steel, other characteristics such as the elastic limit. Of course there were often great variations in the values measured, as indeed there are even today with some materials. In order to ensure a safe design, a lower bound on the test results - a value below which experimental data did not fall - was normally adopted as the 'strength'. Recognising some of the uncertainties associated with design methods based on calculation, stresses under maximum working load conditions were limited to a value equal to the elastic limit divided by a factor of safety. This factor of safety was specified in an apparently arbitrary fashion with values of 4 or 5 being quite typical.

This approach provided the basis of almost all structural design calculations until quite recently, and for some applications is still used today. As understanding of material behaviour has increased and safety factors have become more rationalised, so design strengths have changed. Changes in construction practice, and the development of new, higher strength materials, have necessitated detailed changes in design rules, particularly with regard to buckling behaviour. However the basic approach remained unchanged until quite recently when certain limitations in classical allowable-stress design became apparent. The limitations can be summarised as follows:

i. there is no recognition of the different levels of uncertainty associated with different types of load.

ii. different types of structure may have significantly different factors of safety in terms of collapse, and these differences do not appear in any quantifiable form.

iii. there is no recognition of the ductility and post-yield reserve of strength characteristic of structural steelwork.

The last of these limitations was overcome by the work of Baker [1] and his colleagues in the 1930s when plastic design was developed. This method was based upon ensuring a global factor of safety against collapse, allowing localised 'failure' with a redistribution of bending stresses. A comparison of elastic and plastic design is given by Beal [2].

In recognition of the disadvantages of the allowable stress design method, an alternative approach, known as limit state design has been adopted. Limit state design procedures have now become well established for most structural types and materials. The approach recognises the inevitable variability and uncertainty in quantifying structural performance, including the uncertainties of material characteristics and loading levels. Ideally, each uncertainty is typically treated in a similar manner using statistical techniques to identify typical or characteristic values and the degree of variation to be expected from this norm [3]. It is then possible to derive partial safety factors, one for each aspect of design uncertainty, which are consistent. Thus different load types, for instance, have different factors applied to them. The structure is then examined for a variety of limit states. In that case the structure is designed to fail under factored loading conditions, giving a clearer picture of the margins of safety than was previously the case with allowable stress design.

3.2 Stability

Inadequate strength is not the only cause of collapse. In particular the designer must ensure adequate stability, both of the complete structure (a function of the overall structural form) and of each part of it (dependent on individual member proportions and materials). The latter is generally dealt with by modifying the material strength to account for individual conditions. Overall stability is very much more difficult to quantify and must be carefully considered at the earliest stage of structural design. In this sense structural stability can be defined by the conditions that a structure will neither collapse (completely or partially) due to minor changes, for instance in its form, condition or normal loading, nor be unduly sensitive to accidental actions. Some examples are shown in Figure 4.

In designing for stability the positioning of the main load-bearing elements should provide a clearly defined path for transmitting loads, including wind and seismic actions to the foundations. In considering wind loads on buildings it is important to provide bracing in two orthogonal vertical planes, distributed in such a way as to avoid undue torsional effects, and to recognise the role of the floor structure in transmitting wind loads to these braced areas (Figure 5). The bracing can be provided in a variety of ways, for instance by cross-bracing elements or rigid frame action.

Consideration of accidental actions, such as explosions or impact, is more difficult, but the principle is to limit the extent of any damage caused. Limitation of damage can be achieved by designing for very high loads (not generally appropriate) or providing multiple load paths. Design requires consideration of local damage rendering individual elements of the structure ineffective, and ensuring the remaining structure is able to carry the new distribution of loads, albeit at a lower factor of safety. Alternative strategies are to provide for dissipation of accidental actions, for instance by venting explosions, and to protect the structure, for instance by installing bollards to prevent vehicle impact on columns (Figure 6).

Structural stability must of course be ensured when alterations are to be carried out to existing structures. In all cases stability during execution must be very carefully considered.

3.3 Robustness

In many ways robustness is associated with stability. Construction forms which fulfil the primary function of accommodating normal loading conditions - which are highly idealised for design purposes - may not perform a secondary function when the structure is subject to real loading conditions. For instance the floor of a building is normally expected to transmit wind loads in the horizontal plane to the braced positions. Transmission of wind loads can only be achieved if there is adequate connection between the floor and other parts of the structure and building fabric, and the floor itself is of a suitable form of construction.

4. OTHER DESIGN OBJECTIVES

Although design against collapse is a principal consideration for the structural engineer, there are many other aspects of performance which must be considered. None of these aspects can be quantified and only certain ones will normally apply. However, for a successful solution, the designer must decide which considerations can be ignored, what the most important criteria are in developing the design, and which can be checked simply to ensure satisfactory performance.

4.1 Deformation

The deflection characteristics of a structure are concerned with stiffness rather than strength. Excessive deflections may cause a number of undesirable effects. They include damage to finishes, (particularly where brittle materials such as glass or plaster are used), ponding of water on flat roofs (which can lead to

leaks and even collapse in extreme cases), visual alarm to users and, in extreme cases, changes in the structural behaviour which are sufficient to cause collapse. Perhaps the most common example of deflection effects occurs in columns, which are designed for largely compressive loads but may become subject to significant bending effects when the column deforms in a horizontal plane - the so called P-delta effect.

The normal approach in design is to check that calculated deflections do not exceed allowable levels, which are dependent upon structural type and finishes used. For instance, deflection limits for roof structures are not normally as severe as those for floor structures. In performing these checks it is important to recognise that the total deflection δmax consists of various components, as shown in Figure 7, namely:

δmax = δ1 + δ2 - δ0

where δ1 is the deflection due to permanent loads

δ2 is the deflection due to variable loads

δ0 is the precamber (if any) of the beam in the unloaded state.

In controlling deflections it is often necessary to consider both δmax and δ2, with more severe limits applying in the latter case.

Although the calculated deflections do not necessarily provide an accurate prediction of likely values, they do give a measure of the stiffness of the structure. They are therefore a reasonable guide to structural performance in this respect. With the trend towards longer spans and higher strength materials, design for deflection has become more important in recent years. In many cases this consideration dictates the size of structural elements rather than their resistance. In the case of certain structures, deflection control is of paramount importance. Examples include structures supporting overhead cranes and those housing sensitive equipment. Design for deflection is likely to be the critical condition in such cases.

4.2 Vibration

The vibration characteristics of a structure are, like deflection behaviour, dependent upon stiffness rather than strength. The design principle is to adopt a solution for which the natural frequency of vibration is sufficiently different from any source of excitation, such as machines, to avoid resonance. Longer spans, lighter structures and a reduction in the mass and stiffness of partitions and cladding have all contributed to a general lowering of the natural frequencies for building structures. Cases of human discomfort have been recorded and Eurocode 3 [4] now requires a minimum natural frequency of 3 cycles per second for floors in normal use and 5 cycles per second for dance floors.

Wind excited oscillations may also need to be considered for unusually flexible structures such as very slender, tall buildings, long-span bridges, large roofs, and unusually flexible elements such as light tie rods. These flexible structures should be investigated under dynamic wind loads for vibrations both in-plane and normal to the wind direction, and be examined for gust and vortex induced vibrations. The dynamic characteristics of the structure may be the principal design criterion in such cases.

4.3 Fire Resistance

The provision for safety in the event of fire is dealt with in Group 4B. It is a common requirement that structural integrity is maintained for a specified period to allow building occupants to escape and fire-fighting to be carried out without the danger of structural collapse. For steel structures alternative design strategies can be adopted to achieve this requirement. The traditional approach has been to complete the structural design 'cold' and to provide some form of insulation to the steelwork. This approach can give an expensive solution and alternative methods have now been developed, allowing reductions, and in some cases complete elimination, of fire protection. In order to implement these alternatives in an effective manner, it is important that, at an early stage in the design process, the structural design considers how the fire resistance of the steelwork is to be achieved. Adopting a design solution which may be relatively inefficient in terms of the weight of steel for normal conditions may be more than offset by savings in fire protection (Figure 8).

Buildings close to a site boundary may require special consideration to prevent an outbreak of fire spreading to adjacent sites due to structural collapse. Again quantitative design procedures have been developed for such circumstances [5].

4.4 Fatigue

Where structures, or individual structural elements, are subject to significant fluctuations in stress, fatigue failure can occur after a number of loading cycles at stress levels well below the normal static resistance. The principal factors affecting fatigue behaviour are the range of stresses experienced, the number of cycles of loading and the environment. Structures which need particular consideration in this respect are crane gantry girders, road and rail bridges, and structures subject to repeated cycles from vibrating machinery or wind-induced oscillations. Design guidance is included in Eurocode 3 [4].

4.5 Execution

One of the principal advantages of steelwork is the speed with which execution can proceed. In order to maximise this advantage it may be necessary to adopt a structurally less efficient solution, for instance by using the same profile for all members in a floor construction, even though some floor beams are less highly loaded than others (Figure 9). Temporary propping should be avoided as must late changes in detail which might affect fabrication.

It is important that the structure is not considered in isolation, but rather treated as one part of the complete construction, along with services, cladding and finishes. By adopting a co-ordinated approach to the design, integrating the parts and eliminating or reducing wet trades, speed of execution of the project as a whole can be maximised. A good example of this is the two-way continuous grillage system used for the BMW Headquarters at Bracknell and other projects [6].

The installation of services can have significant implications for speed, cost and detail of construction. In buildings with major service requirements, the cost of the services can be considerably greater than the cost of the structure. In such circumstances it may well be better to sacrifice structural efficiency for ease of accommodating the services. The design of the total floor zone including finishes, structure, fire protection and services also has implications for other aspects of the building construction. The greater the depth of floor construction, the greater the overall height of the building and hence the quantity of external cladding required. In many commercial developments very sophisticated and expensive cladding systems are used. Savings in cladding systems may more than offset the use of shallower, but less efficient, floor construction. Where there is strict planning control of overall building height, it may even be possible to accommodate additional storeys in this way.

4.6 Maintenance

All structures should be inspected and maintained on a regular basis, although some conditions are likely to be more demanding in this respect. For instance, steelwork within a dry, heated interior environment should not suffer from corrosion, whilst a bridge structure in a coastal area will need rigorous maintenance schedules. Some structural forms are easier to maintain than others, and where exposure conditions are severe, ease of inspection and maintenance should be an important criterion. Principal objectives in this context are the avoidance of inaccessible parts, dirt and moisture traps, and the use of rolled or tubular individual sections in preference to truss-like assemblies composed of smaller sections.

5. DESIGN RESPONSIBILITIES

One engineer should be responsible for ensuring that the design and details of all components are compatible and comply with the overall design requirements. This responsibility is most important when different designers or organisations are responsible for individual parts of the structure, such as foundations, superstructure and cladding. It should include an appraisal of the working drawings and other documents to establish, inter alia, that requirements for stability have been incorporated in all elements, and that they can be met during the execution stage.

Effective communication both within the design team and between the designer and constructor before and during execution is essential. Good communication will help to avoid potential design conflicts, for instance when services have to penetrate the structure, and also to promote safe completion of the structure in accordance with the drawings and specification. The constructor may also require information concerning results of site surveys and soil investigations, design loadings, load resistance of members, limits on positions of construction joints, and lifting positions on members to be erected as single pieces. A statement accompanied by sketches detailing any special requirements should be prepared when necessary, e.g. for any unusual design or for any particularly sensitive aspects of the structure or construction. This statement should be made available to the contractor for appropriate action regarding temporary works and execution procedures.

The designer should be made aware of the proposed construction methods, erection procedures, use of plant, and temporary works. The execution programme and sequence of erection should be agreed between the designer and constructor.

Full and effective communication between all parties involved will help not only to promote safe and efficient execution but may also improve design concepts and details. Design should not be seen as an end in itself, but rather as an important part of any construction project.


• There are very many uncertainties associated with structural design. However powerful the tools available, the engineer should always recognise that the design model is no more than an idealisation and simplification of the real condition.

• A quantified approach to structural design can take different forms with a view to providing a framework for satisfactory solutions. The application of design rules should be tempered with common sense and understanding.

• Structural design must consider many aspects of both performance and cost. The most efficient structural solution may not result in the most efficient solution overall if other interdependent aspects of the construction are not considered in a co-ordinated fashion.

7. REFERENCES

[1] Baker, J.F., and Heyman, J. "Plastic Design of Frames 1: Fundamentals", Cambridge University Press, 1969.

[2] Beal, A.N. "What's wrong with load factor design?", Proc. ICE, Vol. 66, 1979.

[3] Armer, G.S.T., and Mayne, J.R. "Modern Structural Design Codes - The case for a more rational format", CIB Journal Building Research and Practice, Vol. 14, No. 4, pp. 212-217, 1986.

[4] Eurocode 3 "Design of Steel Structures" ENV1992-1-1: Part 1: General Rules and Rules for Buildings, CEN, 1992.

[5] Newman, G.J. "The behaviour of portal frames in boundary conditions", Steel Construction Institute.

[6] Brett, P.R. 'An alternative approach to industrial building", The Structural Engineer, Nov. 1982.

Lecture 1B.2.2: Limit State Design Philosophy and Partial Safety Factors

OBJECTIVE/SCOPE

To explain the philosophy of limit state design in the context of Eurocode 3: Design of Steel Structures. To provide information on partial safety factors for loads and resistance and to consider how the particular values can be justified.

RELATED LECTURES




Lecture 2.4: Steel Grades and Qualities

Lecture 2.5: Selection of Steel Quality

SUMMARY

The need for structural idealisations is explained in the context of developing quantitative analysis and design procedures. Alternative ways of introducing safety margins are discussed and the role of design regulations is introduced. The philosophy of limit state design is explained and appropriate values for partial safety factors for loads and strength are discussed. A glossary of terms is included.

1. INTRODUCTION

The fundamental objectives of structural design are to provide a structure which is safe and serviceable to use, economical to build and maintain, and which satisfactorily performs its intended function. All design rules, whatever the philosophy, aim to assist the designer to fulfil these basic requirements. Early design was highly empirical. It was initially based largely upon previous experience, and inevitably involved a considerable number of failures. Physical testing approaches were subsequently developed as a means of proving innovative designs. The first approaches to design based upon calculation methods used elastic theory. They have been used almost exclusively as the basis for quantitative structural design until quite recently. Limit state design is now superseding the previous elastic permissible stress approaches and forms the basis for Eurocode 3 [1] which is concerned with the design of steel structures. In the following sections the principles of limit state design are explained and their implementation within design codes, in particular Eurocode 3, is described.

2. PRINCIPLES OF LIMIT STATE DESIGN

The procedures of limit state design encourage the engineer to examine conditions which may be considered as failure - referred to as limit states. These conditions are classified into ultimate and serviceability limit states. Within each of these classifications, various aspects of the behaviour of the steel structure may need to be checked.

Ultimate limit states concern safety, such as load-carrying resistance and equilibrium, when the structure reaches the point where it is substantially unsafe for its intended purpose. The designer checks to ensure that the maximum resistance of a structure (or element of a structure) is adequate to sustain the maximum actions (loads or deformations) that will be imposed upon it with a reasonable margin of safety. For steelwork design the aspects which must be checked are notably resistance (including yielding, buckling, and transformation into a mechanism) and stability against overturning (Figure 1). In some cases it will also be necessary to consider other possible failure modes such as fracture due to material fatigue and brittle fracture.

Serviceability limit states concern those states at which the structure, although standing, starts to behave in an unsatisfactory fashion due to, say, excessive deformations or vibration (Figure 2). Thus the designer would check to ensure that the structure will fulfil its function satisfactorily when subject to its service, or working, loads.

These aspects of behaviour may need to be checked under different conditions. Eurocode 3 for instance defines three design situations, corresponding to normal use of the structure, transient situations, for example during construction or repair, and accidental situations. Different actions, i.e. various load combinations and other effects such as temperature or settlement, may also need to be considered (Figure 3).

Despite the apparently large number of cases which should be considered, in many cases it will be sufficient to design on the basis of resistance and stability and then to check that the deflection limit will not be exceeded. Other limit states will clearly not apply or may be shown not to govern the design by means of quite simple calculation.

At its most basic level limit state design simply provides a framework within which explicit and separate consideration is given to a number of distinct performance requirements. It need not necessarily imply the automatic use of statistical and probabilistic concepts, partial safety factors, etc., nor of plastic design, ultimate load design, etc. Rather it is a formal procedure which recognises the inherent variability of loads, materials, construction practices, approximations made in design, etc., and attempts to take these into account in such a way that the probability of the structure becoming unfit for use is suitably small. The concept of variability is important because the steelwork designer must accept that, in performing his design calculations, he is using quantities which are not absolutely fixed or deterministic. Examples include values for loadings and the yield stress of steel which, although much less variable than the properties of some other structural materials, is known to exhibit a certain scatter (Figure 4). Account must be taken of these variations in order to ensure that the effects of loading do not exceed the resistance of the structure to collapse. This approach is represented schematically in Figure 5 which shows hypothetical frequency distribution curves for the effect of loads on a structural element and its strength

or resistance. Where the two curves overlap, shown by the shaded area, the effect of the loads is greater than the resistance of the element, and the element will fail.

Proper consideration of each of the limits eliminates the inconsistencies of attempting to control deflection by limiting stresses or of avoiding yield at working load by modifying the design basis (formula, mathematical model, etc.) for an ultimate resistance determination.

The procedure of limit state design can therefore be summarised as follows:

• define relevant limit states at which the structural behaviour is to be checked. • for each limit state determine appropriate actions to be considered. • using appropriate structural models for design, and taking account of the inevitable variability of

parameters such as material properties and geometrical data, verify that none of the relevant limit states is exceeded.

3. ACTIONS

An action on a structure may be a force or an imposed deformation, such as that due to temperature or settlement. Actions are referred to as direct and indirect actions respectively in Eurocode 3.

Actions may be permanent, e.g. self-weight of the structure and permanent fixtures and finishes, variable, e.g. imposed, wind and snow loads, or accidental, e.g. explosions and impact (Figure 6). For earthquake actions, see Lectures 17 and Eurocode 8 [2]. Eurocode 1 [3] represents these by the symbols G, Q and A respectively, together with a subscript - k or d to denote characteristic or design load values respectively. An action may also be classified as fixed or free depending upon whether or not it acts in a fixed position relative to the structure.

3.1 Characteristic Values of Actions (Gk, Qk and Ak)

The actual loadings applied to a structure can seldom be defined with precision; liquid retaining structures may provide exceptions. To design a structure for the maximum combination of loads which could conceivably be applied would in many instances be unreasonable. A more realistic approach is to design the structure for 'characteristic loads', i.e. those which are deemed to have just acceptable probability of

not being exceeded during the lifetime of the structure. The term 'characteristic load' normally refers to a load of such magnitude that statistically only a small probability, referred to as the fractile, exists of it being exceeded.

Imposed loadings are open to considerable variability and idealisation, typically being related to the type of occupancy and represented as a uniform load intensity (Figure 7). Dead loads are less variable although there is evidence that variations arising in execution and errors can be substantial, particularly in the case of in-situ concrete and finishes such as tarmac surfacing on road bridges.

Loadings due to snow, wind, etc. are highly variable. Considerable statistical data on their incidence have been collated. Consequently it is possible to predict with some degree of certainty the risk that these environmental loads will exceed a specified severity for a particular location.

3.2 Design Values of Actions (Gd, Qd and Ad)

The design value of an action is its characteristic value multiplied by an appropriate partial safety factor. The actual values of the partial factors to be used depend upon the design situation (normal, transient or accidental), the limit state and the particular combination of actions being considered. Corresponding values for the design effects of actions, such as internal forces and moments, stresses and deflections, are determined from the design values of the actions, geometrical data and material properties.

4. MATERIAL PROPERTIES

Variability of loading is only one aspect of uncertainty relating to structural behaviour. Another important one is the variability of the structural material which is reflected in variations in strength of the components of the structure. Again, the variability is formally accounted for by applying appropriate partial safety factors to characteristic values. For structural steel, the most important property in this context is the yield strength.

4.1 Characteristic Values of Material Properties

The characteristic yield strength is normally defined as that value below which only a small proportion of all values would be expected to fall. Theoretically this can only be calculated from reliable statistical data.

In the case of steel, for practical reasons a nominal value, corresponding typically to the specified minimum yield strength, is generally used as the characteristic value for structural design purposes. This is the case in Eurocode 3 which tabulates nominal values of yield strength for different grades of steel.

4.2 Design Values of Material Properties

The design value for the strength of steel is defined as the characteristic value divided by the appropriate partial safety factor. Other material properties, notably modulus of elasticity, shear modulus, Poisson's ratio, coefficient of linear thermal expansion and density, are much less variable than strength and their design values are typically quoted as deterministic.

In addition to the quantified values used directly in structural design, certain other material properties are normally specified to ensure the validity of the design procedures included within codified rules. For instance Eurocode 3 stipulates minimum requirements for the ratio of ultimate to yield strength, elongation at failure and ultimate strain if plastic analysis is to be used [1].

5. GEOMETRICAL DATA

Geometrical data are generally represented by their nominal values. They are the values to be used for design purposes. The variability, for instance in cross-section dimensions, is accounted for in partial safety factors applied elsewhere. Other imperfections such as lack of verticality, lack of straightness, lack of fit and unavoidable minor eccentricities present in practical connections should be allowed for. They may influence the global structural analysis, the analysis of the bracing system, or the design of individual structural elements and are generally accounted for in the design rules themselves.

6. PARTIAL SAFETY FACTORS

Instead of the traditional single factor of safety used in permissible stress design, limit state design provides for a number of partial safety factors to relate the characteristic values of loads and strength to design values. ISO Standard 2394 [4] suggests the use of seven partial safety factors but these are often combined to simplify design procedures. This is the case in the Eurocodes [1,3] which include factors for actions and resistance. Further details are given in the Appendix.

In principle, the magnitude of a partial safety factor should be related to the degree of uncertainty or variability of a particular quantity (action or material property) determined statistically. In practice, whilst this appears to be the case, the actual values of the partial safety factors used incorporate significant elements of the global safety factor and do not represent a rigorous probabilistic treatment of the uncertainties [5-8].

In essence the characteristic actions (Fk) are multiplied by the partial safety factors on loads (γF) to obtain the design loads (Fd), that is:

Fd = γf Fk

The effects of the application of the design loads to the structure, i.e. bending moment, shear force, etc. are termed the 'design effects' Ed.

The design resistance Rd is obtained by dividing the characteristic strengths Rk by the partial safety factors on material γM, modified as appropriate to take account of other considerations such as buckling. For a satisfactory design the design resistance should be greater than the 'design effect'.

7. ULTIMATE LIMIT STATE

The following conditions may need to be verified under appropriate design actions:

a. Ed,dst ≤ Ed,stb

where Ed,dst and Ed,stb are the design effects of destabilising and stabilising actions respectively. This is the ultimate limit state of static equilibrium.

b. Ed ≤ Rd

where Ed and Rd are the internal action and resistance respectively. In this context it may be necessary to check several aspects of an element's resistance. These aspects might include the resistance of the cross-section (as a check on local buckling and yielding), and resistance to various forms of buckling (such as overall buckling in compression, lateral-torsional buckling and shear buckling of webs), as well as a check that the structure does not transform into a mechanism.

c. no part of the structure becomes unstable due to second order effects.

d. the limit state of rupture is not induced by fatigue.

8. SERVICEABILITY LIMIT STATE

The serviceability limit state is generally concerned with ensuring that deflections are not excessive under normal conditions of use. In some cases it may also be necessary to ensure that the structure is not subject to excessive vibrations. Cases where this is particularly important include structures exposed to significant dynamic forces or those accommodating sensitive equipment. Both deflection and vibration are associated with the stiffness rather than strength of the structure.

8.1 Deflections

At the serviceability limit state, the calculated deflection of a member or of a structure is seldom meaningful in itself since the design assumptions are rarely realised because, for example:

• the actual load may be quite unlike the assumed design load. • beams are seldom "simply supported" or "fixed" and in reality a beam is usually in some

intermediate condition. • composite action may occur.

The calculated deflection is, however, valuable as an index of the stiffness of a member or structure, i.e. to assess whether adequate provision is made in relation to the limit state of deflection or local damage. For this purpose, sophisticated analytical methods are seldom justified. Whatever methods are adopted to assess the resistance and stability of a member or structure, calculations of deflection should relate to the structure of the elastic state. Thus, when analysis to check compliance with the strength limit is based on

rigid-elastic or elastic-plastic concepts, the structural behaviour in the elastic phase must also be considered.

Calculated deflections should be compared with specified maximum values, which will depend upon circumstances. Eurocode 3 [1] for instance tabulates limiting vertical deflections for beams in six categories as follows:

• roofs generally. • roofs frequently carrying personnel other than for maintenance. • floors generally. • floors and roofs supporting plaster or other brittle finish or non-flexible partitions. • floors supporting columns (unless the deflection has been included in the global analysis for the

ultimate limit state). • situations in which the deflection can impair the appearance of the building.

In determining the deflection it may be necessary to consider the effects of precamber, permanent loads and variable loads separately. The design should also consider the implications of the deflection values calculated. For roofs, for instance, regardless of the limits specified in design rules, there is a clear need to maintain a minimum slope for run-off. More stringent limits may need therefore to be considered for nearly flat roof structures.

8.2 Dynamic Effects

The dynamic effects to be considered at the serviceability limit state are vibration caused by machinery and self-induced vibrations, e.g. vortex shedding. Resonance can be avoided by ensuring that the natural frequencies of the structure (or any part of it) are sufficiently different from those of the excitation source. The oscillation and vibration of structures on which the public can walk should be limited to avoid significant discomfort to the users. This situation can be checked by performing a dynamic analysis and limiting the lowest natural frequency of the floor. Eurocode 3 recommends a lower limit of 3 cycles per second for floors over which people walk regularly, with a more severe limit of 5 cycles per second for floors used for dancing or jumping, such as gymnasia or dance halls [1]. An alternative method is to ensure adequate stiffness by limiting deflections to appropriate values.

9. STRUCTURAL DESIGN MODELS

No structural theory, whether elastic or plastic, can predict the load-carrying resistance of a structure in all circumstances and for all types of construction. The design of individual members and connections entails the use of an appropriate structural theory to check the mode of failure; sometimes alternative types of failure may need to be checked and these may require different types of analysis. For example, bending failure by general yielding can only occur when the plastic moment is attained; however bending failure is only possible if failure does not occur at a lower load level by either local or overall buckling.

Serviceability limit states are concerned with the performance of the structure under service loading conditions. The behaviour should therefore be checked on the basis of an elastic analysis, regardless of the model used for the ultimate limit state design.


• Limit state design procedures require formal examination of different conditions which might lead to collapse or inadequate performance.

• The effect of various actions is compared with the corresponding resistance of the structure under defined failure criteria (limit states).

• The most important failure critera are the ultimate limit state (collapse) and the serviceability limit state of deflection.

• In checking each limit state, appropriate design models must be used to provide an accurate model of the corresponding structural behaviour.

• Separate partial safety factors are introduced for loading and material. These factors are variable quantities and the precise values to be used in design reflect the degree of variability in the action or resistance to be factored.

• Different combinations of action may also require different values of safety factor. • This flexible approach helps provide a more consistent level of safety compared with other design

approaches.

11. GLOSSARY

A limit state is a condition beyond which the structure no longer satisfies the design performance requirements.

The ultimate limit state is a state associated with collapse and denotes inability to sustain increased load.

The serviceability limit state is a state beyond which specified service requirements are no longer met. It denotes loss of utility and/or a requirement for remedial action.

Characteristic loads (Gk, Qk, Ak) are those loads which have an acceptably small probability of not being exceeded during the lifetime of the structure.

The characteristic strength (fy) of a material is the specified strength below which not more than a small percentage (typically 5%) of the results of tests may be expected to fall.

Partial safety factors (γ G, γ Q, γ M) are the factors applied to the characteristic loads, strengths, and properties of materials to take account of the probability of the loads being exceeded and the assessed design strength not being reached.

The design (or factored) load (Gd, Qd, Ad) is the characteristic load multiplied by the relevant partial safety factor.

The design strength is the characteristic strength divided by the appropriate partial safety factor for the material.

12. REFERENCES

[1] Eurocode 3: "Design of Steel Structures" ENV 1993-1-1: Part 1.1: General Rules and Rules for Buildings, CEN, 1992.

[2] Eurocode 8: "Structures in Seismic Regions-Design", CEN (in preparation).

[3] Eurocode 1: "Basis of Design and Actions on Structures" CEN (in preparation).

[4] ISO 2394, General Principles for the Verification of the Safety of Structures, International Standards Organisation, 1973.

[5] Rationalisation of Safety and Serviceability Factors in Structural Codes, CIRIA Report 63, London, 1972.

[6] Allen, D. E., "Limit States Design - A Probabilistic Study", Canadian Journal of Civil Engineers, March 1975.

[7] Augusti, G., Baratta, A., and Casciati, F., "Probabilistic Methods in Structural Engineering", Chapman and Hall, London 1984.

[8] Armer, G. S. T., and Mayne, J. R, "Modern Structural Design Codes - The Case for a More Rational Format", CIB Journal Building Research and Practice, Vol. 14, No. 4, pp 212-217, 1986.

13. ADDITIONAL READING

1. Pugsley, A., "The Safety of Structures", Edward Arnold, London 1966.

2. Thoft-Christensen, P., and Baker, M. J., "Structural Reliability Theory and its Application", Springer-Verlag, 1982.

3. "The Steel Skeleton", Cambridge University Press, Vol 1 1960, Vol II 1965.

4. Blockley, D., "The Nature of Structural Design and Safety", Ellis Horwood, Chichester, 1980.

5. Fukumoto, Y., Itoh, Y. and Kubo, M., "Strength Variation of Laterally Unsupported Steel Beams", ASCE, Vol 106, ST1, 1980.

6. ISO 8930: General Principles on Reliability of Structures - List of Equivalent Terms, 1987.

APPENDIX - PARTIAL SAFETY FACTORS

Partial safety factors for actions

Eurocodes 1 and 3 define three partial safety factors as follows:

γG permanent actions

γQ variable actions

γA accidental actions

Two values are specified for γG. These are γG,sup and γG,inf representing 'upper' and 'lower' values respectively. Where permanent actions have an adverse effect on the design condition under consideration, the partial safety factor should be the upper value. However, where the effect of a permanent action is favourable (for instance in the case of loads applied to a cantilever when considering the design of the adjacent span), the lower value for the partial safety factor should be used, see Figure 8.

The treatment of load combinations is quite sophisticated, and involves the definition of 'representative' values, determined by applying a further factor to the design loads, depending upon the particular combination considered. However, simplified procedures are generally permitted. They are outlined below. Note that the values of partial safety factors are indicative only. Although they are specified in Eurocode 3, their precise value may be adjusted by individual countries for use within the country.

Load combinations for the ultimate limit state

Either, all permanent loads plus one variable load, all factored, i.e:

Σ γG Gki + γQ Qk1

where γG and γQ are taken as 1,35 and 1,5 respectively,

or, all permanent loads plus all variable loads, all factored, i.e:

Σ γG Gki + Σ γQ Qki

where γG and γQ are both taken as 1,35.

These values recognise the reduced probability of more than one variable load existing simultaneously. For instance, although a structure may on occasions be subject to its maximum wind load, it is much less likely that it will be exposed to a combination of maximum wind and imposed loads.

Load combinations for the serviceability limit state

Either, all permanent loads plus one variable load are considered. In each case the partial safety factor is unity, i.e. the loads are unfactored characteristic values:

Σ Gki + Qk1

or, all permanent loads (partial safety factor unity) plus all variable loads (with a partial safety factor of 0,9), i.e:

Σ Gki + 0,9 Σ Qki

Where simplified compliance rules are provided for serviceability, there is no need to perform detailed calculations with different load combinations.

Partial safety factors for material

Alternative partial safety factors for material are specified as follows:

γM0 = 1,1 for consideration of resistance of Class 1, 2 or 3 cross-section.

γM2 = 1,1 for consideration of resistance of Class 4 cross-section and resistance to buckling.

γM2 = 1,25 for resistance consideration of cross-section at holes


OBJECTIVE/SCOPE:

To provide an introduction to the sources of loads on structures and how loads can be quantified for the purpose of structural design.

RELATED LECTURES:


SUMMARY:

Various types of loads (dead, imposed and environmental) and their classification as permanent, transient or accidental within Eurocode 1: Basis of Design and Actions on Structures, is considered. Calculations for dead loads on the basis of material densities and component sizes are explained. Means of estimating imposed loads based upon usage and the implications of change of use are discussed. Loads due to snow, temperature and seismic effects are considered briefly. The statistical treatment of wind and wave loads, and their dependence upon wind speed and wave height respectively, are described. The importance of load characteristics, other than simply their magnitude, is considered. These characteristics include fatigue, dynamic and aerodynamic effects. Simplified treatments for dynamic loads are described.

1. INTRODUCTION

Structures are subject directly to loads from various sources. These loads are referred to as direct actions and include gravity and environmental effects, such as wind and snow. In addition deformations may be imposed on a structure, for instance due to settlement or thermal expansion. These 'loads' are indirect actions. In applying any quantitative approach to structural analysis, the magnitudes of the actions need to be identified. Furthermore, if the structure is to perform satisfactorily throughout its design life, the nature of the loads should be understood and appropriate measures taken to avoid problems of, for instance, fatigue or vibration.

The magnitude of loads cannot be determined precisely. In some cases, for instance in considering loads due to the self-weight of the structure, it might be thought that values can be calculated fairly accurately. In other cases, such as wind loads, it is only possible to estimate likely levels of load. The estimate can be based on observation of previous conditions and applying a probabilistic approach to predict maximum effects which might occur within the design life of the structure. (In fact, the extensive wind records which are now available mean that wind loads can often be predicted with greater accuracy than self-weight). Loads associated with the use of the structure can only be estimated based on the nature of usage. Insufficient data is available in most cases for a fully statistical approach and nominal values are therefore assigned. In addition, problems of change of use and fashion can occur.

In analysing structures it is rare to consider all loadings acting simultaneously. This approach may be because the most severe condition for parts of the structure occurs when some other combination of load is considered. Alternatively it may be that the possibility of such a condition actually occurring is extremely small. However, the risk of coexistence of apparently unrelated loads may be greater than is first imagined. Correlations can be produced from unexpected sources or from coincidences which, although physically unconnected, are temporarily connected. For example, floor and wind loads would normally be considered as unrelated. However, in hurricane areas residents on the coast might be

expected to move their ground floor contents to upper floors if a hurricane warning, with associated tidal surge, were given. This circumstance could very easily produce extreme floor loads in combination with extreme wind loads. This case may be a very special one but there are others. The risk of fire may not be considered correlated with high wind loads, yet in many parts of the world high winds are more likely in winter, which is also the period of greatest fire risk.

For these reasons it is convenient to consider loads under various categories. The categories can then be ascribed different safety factors and applied in various combinations as required. Traditionally, loadings have been classified as dead, superimposed and environmental loads. These classes include a wide range of gravity effects, seismic action, pressures due to retained material or liquids, temperature induced movement, and, for marine structures, water movement. The Eurocodes on actions and steelwork design [1, 2] classify loads and other actions as permanent, variable and accidental. These classes of action will be considered in more detail in the following Sections.

In limit state design, characteristic values of actions are used as the basis of all calculations. They are values which statistically have only a small probability of being exceeded during the life of the structure. To provide a margin of safety, particularly against collapse, partial safety factors are applied to these characteristic values to obtain design quantities. In principle, different partial safety factors can be applied depending on the degree of uncertainty or variability of a particular type of action. In practice, whilst this appears to be the case, the actual values of partial safety factors used incorporate significant elements of the global safety factor and do not represent a rigorous probabilistic treatment of the uncertainties of the actions.

2. PERMANENT ACTIONS

Permanent actions, as the name implies, are always present and must be considered in all cases. They comprise what are traditionally referred to as dead loads, but may also include permanent imposed loads due, for instance, to machinery or stored material.

2.1 Dead Loads

Dead loads are gravity loads due to the self weight of the structure and any fixtures or finishes attached to it (Figure 1). Their magnitudes can be estimated with reasonable confidence based on prescribed dimensions and a knowledge of material density. Even so, variations due to constructional tolerances and natural variations in materials, will exist. Furthermore, fixtures, fittings and finishes may be replaced or modified during the life of the structure. This possibility has been recognised in calculating loads on bridge decks, for which a separate load category of 'superimposed dead load' is included to allow for surfacing which is likely to be replaced a number of times during the life of the bridge. For this situation there is consequently a much greater potential for variability than for other dead loads.

A similar condition exists within certain types of building with respect to partitions (Figure 2). Where the position of walls is predetermined their weight can simply be included as a dead load. For more speculative development, internal partitions will be the responsibility of the client and their layout is likely to change many times during the life of the building. An allowance, as an equivalent uniformly distributed load, is therefore normally made.

Schedules of densities for common building materials are listed in Eurocode 1 [1] and manufacturers of proprietary products, such as cladding, blockwork, raised floors, etc. provide information on weights. Together with specified dimensions, these data enable dead loads to be calculated. Where dead loads are not strictly evenly distributed over a plan area, such as timber floor joists located at discrete intervals, they are often represented as an equivalent uniformly distributed load for convenience in design calculations. As long as the equivalent magnitude is determined in a rational manner, any differences between this simplified approach and a more rigorous analysis taking account of the actual location of the joists will be negligible.

To determine dead loads, consider, for example, the case of a floor consisting of a 150mm thick reinforced concrete slab with 50mm lightweight screed and a 15mm plaster soffit. Details are shown in Figure 3 together with densities for each material. The total dead load per square metre of floor plan can be calculated as follows:

lightweight screed 15 x 0,05 = 0,75 kN/m2

rc slab 24 x 0,15 = 3,60

plaster 12 x 0,015 = 0,18

total dead load = 4,53 kN/m2

In addition an allowance would normally be made for any services or fittings (electric lighting, pipework, etc.) fitted to the underside of the slab or located within the screed or under a raised floor (Figure 4). This case is another where an equivalent uniformly distributed load is used to represent load sources distributed in an uneven manner. A value between 0,1 and 0,3 kN/m2 is normally adequate to cover such installations.

The weight of walls can be treated in a similar manner to floors by considering the various component parts and summing the weights per square metre on elevation. For example, consider a cavity wall consisting of a tile-hung brick outer leaf (100mm thick) and a plastered blockwork inner leaf (150mm thick) as shown in cross-section in Figure 5.

The total dead load is determined as follows:

tiles 0,6 kN/m2

brickwork 2,1

blockwork 1,4

plaster 0,2

total dead load of wall 4,3 kN/m2

By multiplying this value by the height of the wall, the load intensity as a line load on the supporting structure can be determined.

Loads due to internal lightweight stud or blockwork partitions cannot normally be treated in such a rigorous manner since their location is often not known at the design stage and in any case may change during the life of the building. Instead an allowance is made within the assessment of imposed loads which is described under variable actions.

3. VARIABLE ACTIONS

Variable actions comprise loads which are not always acting but may exist at various times during the normal use of the structure. They include loads due to the occupation of a building and traffic on bridges (imposed loads), snow and wind loads (environmental loads), and temperature effects (Figure 6). They do not include accidental conditions such as fire, explosion or impact.

3.1 Imposed Loads

Imposed loads - sometimes referred to as "superimposed", "super" or "live" loads - are those loads due directly to the use of the structure. For buildings, they are concerned with the occupancy by people, furniture, equipment, etc. For bridges they are due to traffic, whether pedestrian or vehicular.

Clearly these conditions will be almost constantly changing and are rather more difficult to quantify than dead loads. For buildings, the approach has therefore been to relate imposed load levels to occupancy, and to base them on observation and sensible deduction. Eurocode 1: Basis of Design and Actions on Structures [1] distinguishes between four classes of loaded floor area as follows:

• areas of dwellings, offices, etc. • garage and traffic areas. • areas for storage, production machinery and filing. • areas serving as escape routes.

The first class is further subdivided into four categories according to their specific use. They are residential (including hospital wards, hotel bedrooms etc.), public premises (such as offices, hotels, hospitals, schools, leisure centres etc.), public premises susceptible to overcrowding (including assembly halls, conference rooms, theatres, shopping areas and exhibition rooms), and public premises susceptible to overcrowding and accumulation of goods (including areas in warehouses and department stores).

The characteristic values of the imposed loads for these different categories are given in Table 1. Thus domestic residences attract a lower imposed load than office accommodation; areas of public assembly, where large numbers of people could gather at any one time, are prescribed a high superimposed load. Storage areas must be particularly carefully considered and Eurocode 1 includes details of densities for a range of stored materials. Some of these, such as steel strip, will generate high loads, but even apparently innocuous conditions, such as filing stores, can experience very high loading levels. Escape routes must be designed for relatively high imposed loads.

Although such loads are used in limit state design in a semi-probabilistic way and are referred to as characteristic values (implying a statistical basis for their derivation) little data is available. A proper statistical analysis is not therefore possible and values specified are nominal quantities. One study which was conducted into office accommodation in the UK [4] revealed a wide variation in actual load levels for similar building occupancies. In all cases the load levels measured were considerably less than the characteristic values specified for the structural design. However, this observation must be viewed with some caution since design must allow for extreme conditions, misuse and panic situations.

Note that, although imposed loading will rarely be evenly distributed, a uniform distribution of load intensity is normally assumed (Figure 7).

3.2 Permitted Reductions in Imposed Load

The nominal values of imposed load associated with different classifications of building occupancy and use represent extreme conditions. In many cases the probability of such conditions existing simultaneously throughout a building is remote. In recognition of this remote possibility some reductions in imposed load intensity may be permitted. Reduction applies particularly to columns in multi-storey buildings where it increases with the number of floors supported by a particular length of column. Typical reductions range from 10% to 30% and apply to imposed loads only. No reductions are permitted in dead load or for certain types of imposed load - notably in the case of storage areas, crane loads, and loads explicitly allowed for such as those due to machinery or due to people in public premises susceptible to overcrowding.

3.3 Superimposed Bridge Loads

In practice a highway bridge is loaded in a very complex way by vehicles of varying sizes and groupings. In order to simplify the design process this real loading is typically simulated by two basic imposed loads - a uniformly distributed load and a knife edge load - representing an extreme condition of normal usage (Figure 8). The design is then checked for a further load arrangement representing the passage of an abnormal load. The magnitudes of all these loads are generally related to the road classification, the highway authority's requirements and the loaded length of the bridge.

For vehicular traffic within buildings, lightweight conditions (less than 16 tonnes) can be dealt with in categories such as cars, light vehicles and medium vehicles. For heavier traffic, highway loading must be considered.

Railway bridge design must take account of static loading and forces associated with the movement of vehicles. As for highway bridges, two models of loading are specified for consideration as separate load cases. They represent ordinary traffic on mainline railways and, where appropriate, abnormal heavy loads. They are expressed as static loads due to stationary vehicles and are factored to allow for dynamic effects associated with train speeds up to 300km/h. Eurocode 1 also gives guidance on the distribution of loads and their effects and specifies horizontal forces due to vehicle motion. Centrifugal forces associated with the movement around curves, lateral forces due to oscillation of vehicles (nosing) and longitudinal forces due to traction and braking are included.

Other aspects of bridge loading which need to be considered include accidental loads and the possibility of premature failure due to fatigue under traffic loading.

3.4 Crane Loads

For buildings fitted with travelling overhead cranes, the loads due to the crane itself and the lifted load are considered separately. The self weight of the crane installation is generally readily available from the manufacturer, and the load lifted corresponds to the maximum lifting capacity of the crane. When a load is lifted from rest, there is an associated acceleration in the vertical direction. In the same way that gravity loads are equal to mass multiplied by the acceleration due to gravity, so the lifting movement causes an additional force. If the load is lifted very gently - that is with little acceleration - this force will be very small, but a sudden snatch, i.e. a rapid rate of acceleration, would result in a significant force. This force is of course in addition to the normal force due to gravity, and is generally allowed for by factoring the normal static crane loads.

Movements of the crane, both along the length and across the width of the building, are also associated with accelerations and retardations, this time in the horizontal plane. The associated horizontal forces must be taken into account in the design of the supporting structure. The magnitude of the forces will depend, as before, on the rates of acceleration. The normal procedure is to calculate the magnitudes on the basis of a proportion of the vertical wheel load.

The approach yields an equivalent static force which can be used in designing the structure for strength. However, the nature of crane loads must also be recognised. The possibility of premature failure due to fatigue under the cyclic loading conditions should be considered.

3.5 Environmental Loads

Environmental loads are clearly variable actions. For bridges and buildings the most important environmental loads are those due to snow and wind. For marine structures, particularly offshore installations such as oil platforms, loads due to water movements are often dominant. The action of waves generally represents the most severe condition. In certain geographical locations, the effects of earthquakes must be included in the structural analysis. All of these loads from environmental sources are beyond the control of man. It has therefore been recognised that a statistical approach must be adopted in order to quantify corresponding design loads.

The approach is based on the 'return period' which is a length of time to which recorded environmental data, such as wind speeds, snowfall or wave heights, is related. If records are only available over a relatively short period, data for the 'return period' may be predicted. The most severe condition on average over the return period then represents the design value. For a return period of 100 years, for example, it is referred to as the 1 in 100 year wind speed or wave height, etc. The return period normally corresponds to the design life of the structure. Clearly there is a degree of uncertainty about the process of predicting the most severe conditions likely to be encountered. Further simplifications are implicit in translating measured environmental data such as wind speeds or wave heights into loads.

3.6 Wind Loads

Wind forces fluctuate with time but for many structures the dynamic effect is small and the wind load can be treated using normal static methods. Such structures are defined as 'rigid' and Eurocode 1 [1] provides guidance on this classification. For slender structures the dynamic effect may be significant. Such structures are classified as 'flexible' structures and their dynamic behaviour must be taken into account.

The most important parameter in quantifying wind loads is the wind speed. The basis for design is the maximum wind speed (gust) predicted for the design life of the structure. Factors which influence its magnitude are:

• geographical location; wind speeds are statistically greater in certain regions than others. For many areas considerable statistical data is now available and basic wind speeds are published usually in the form of isopleths (Figure 9) which are lines of equal basic wind speed superimposed on a map. The basic wind speed is referred to in Eurocode 1 [1] as the reference wind speed and corresponds to the mean velocity at 10m above flat open country averaged over a period of 10 minutes with a return period of 50 years.

• physical location; winds gust to higher speeds in exposed locations such as coasts than in more sheltered places such as city centres (Figure 10), because of varying surface roughness which reduces the wind speed at ground level. This variation is taken into account by a roughness coefficient which is related to the roughness of the terrain and the height above ground level.

• topography; the particular features of a site in relation to hills or escarpments are taken into account by a topography coefficient.

• building dimensions; height is important in particular because wind speeds increase with height above ground level (Figure 11).

• the mean wind velocity is determined by the reference wind velocity factored to account for the building height, ground roughness and topography. The wind pressure is proportional to the square of the mean wind speed. In addition the following parameters are important:

• structural shape; it is important to recognise that wind loads are not simply a frontal pressure applied to the facade of a structure but are the result of a complex pressure distribution on all faces due to the movement of air around the whole structure. The distribution is further

complicated by adjacent structures and natural obstructions/variations such as hills, valleys, woodland which may all influence the pattern of air movement and associated pressure distribution.

• roof pitch; this parameter is really a special aspect of structural shape. It is worth noting that roofs with a very shallow pitch may be subject to uplift or suction, whilst steeper roofs - say greater than about 20° - are likely to be subject to a downwards pressure (Figure 12).

• wind direction; pressure distributions will change for different wind directions (Figure 13). • gust response factor; this factor is used to take into account the reduction of the spatial average of

the wind pressure with increasing area due to the non-coincidence of maximum local pressures acting on the external surface of the structure. Thus small parts of a building, such as cladding units and their fixings, must be designed for higher wind pressures than the whole structure. The gust response factor is related to an equivalent height, which corresponds approximately to the centroid of the net wind force on a structure.

Tabulated procedures enable the above parameters to be accounted for firstly in calculating the design wind speed, and secondly in translating that wind speed into a system of forces on the structure. These equivalent static forces can then be used in the analysis and resistance design of the structure, as a whole. However, certain additional features of wind should also be taken into account:

• local pressures, particularly at corners and around obstructions in an otherwise 'smooth' surface, can be significantly higher than the general level (Figure 14). High local pressures particularly affect cladding and fixing details, but can also be a consideration for structural elements in these areas.

• structures sensitive to wind should be given a more sophisticated treatment. It might involve wind tunnel testing and include the influence of surrounding buildings. Structures which might need to be treated in this way include high-rise buildings, long or slender bridges, masts and towers.

• aerodynamic instability may be a consideration for certain types of structure or component, for example chimneys and masts. Vortex shedding can normally be avoided by the use of strakes (Figure 15). Galloping may be a problem in cables.

3.7 Snow Loads

Loads due to snow have traditionally been treated by specifying a single load intensity, with possible reductions for steep roof slopes. This approach takes no account of such aspects as the increased snowfall at higher altitudes or of locally higher loads due to drifting. Cases of complete or partial collapse due to snow load are not unknown [5]. A more rational approach is to use a snow map giving basic snow load intensities for a specified altitude and return period similar to the treatment for basic wind speeds (Figure 16). Corrections for different altitudes or design life can then be applied as shown in Table 2. At present the European snow map is provisional and further work is under way to acquire more data.

Allowance for different roof configurations can be dealt with by means of a shape coefficient. It provides for conditions such as accumulations of snow behind parapets, in valleys and at abrupt changes of roof height (Figure 17). In addition to snow falling in calm conditions, it may be necessary to consider the effects of wind. Wind may cause a redistribution of snow, and in some cases its partial removal from roofs. Any changes in snow distribution on roofs due to excessive heat loss through part of the roof or snow clearing operations should be accounted for if such loading patterns are critical. Eurocode 1 [1] does not cover additional wind loads due to the presence of snow or the accretion of ice, nor loads in areas where snow is present throughout the year.

3.8 Wave Loading

For offshore structures in deep and hostile waters, wave loads can be particularly severe. The loads arise due to movement of water associated with wave action. These movements can be described mathematically to relate forces to physical wave characteristics such as height and wavelength.

The treatment is therefore similar to wind loads in that these physical characteristics are predicted and corresponding forces on the particular structural arrangement then calculated. These calculation procedures are, however, very complicated and must realistically be performed on a computer.

3.9 Temperature Effects

Exposed structures such as bridges may be subject to significant temperature variation which must be taken into account in the design. If it is not provided for in terms of allowing for expansion, significant forces may develop and must be included in the design calculations. In addition, differential temperatures, e.g. between the concrete deck and steel girders of a composite bridge, can induce a stress distribution which must be considered by the designer.

3.10 Retained Material

Structures for retaining and containing material (granular or liquid) will be subject to a lateral pressure. For liquids it is simply the hydrostatic pressure. For granular material a similar approach can be adopted, but with a reduction in pressure depending on the ability of the material to maintain a stable slope - this is the Rankine approach. Ponding of water on flat roofs should be avoided by ensuring adequate falls (1:60 or more) to gutters.

3.11 Seismic Loads

In some parts of the world earthquakes are a very important design consideration. Seismic actions on structures are due to strong ground motion. They are a function of the ground motion itself and of the dynamic characteristics of the structure.

Strong ground motion can be measured by one of its parameters, the maximum ground acceleration being the parameter most usually adopted for engineering purposes. These parameters are expressed on a probabilistic basis, i.e. they are associated with a certain probability of occurrence or to a return period, in conjunction with the life period of the structure [3].

3.12 Accidental Loads

Accidental actions may occur as a result of accidental situations. The situations include fire, impact or explosion. It is very difficult to quantify these effects. In many cases it may be preferable to avoid the problem, for instance by providing crash barriers to avoid collision from vehicles or roof vents to dissipate pressures from explosions.

Where structures such as crash barriers for vehicles and crowds must be designed for 'impact' the loading is treated as an equivalent static load.


• There are many sources of structural loads, notably dead loads, those due to the use of the structure and environmental effects such as wind, earthquake, snow and temperature. The loads must be quantified for the purpose of structural design. Dead loads can be calculated. Imposed loads can only be related to type of use through observation on other similar structures. Environmental loads are based on a statistical treatment of recorded data.

• Calculated or prescribed values of loads are factored to provide an adequate margin of safety. The nature, as well as the magnitude, of the loads must be recognised, particularly in terms of dynamic and fatigue behaviour.

5. REFERENCES

[1] Eurocode 1: Basis of Design and Actions on Structures, CEN (in preparation).

[2] Eurocode 3: Design of Steel Structures: ENV 1993-1-1: Part 1.1, General principles and rules for buildings, CEN, 1992.

[3] Eurocode 8: Structures in Seismic Regions - Design, CEN (in preparation).

[4] Floor Loadings in Office Buildings - the Results of a Survey, BRE Current Paper 3/71, Building Research Establishment, Watford, 1971.

[5] Design Practice and Snow Loading - Lessons from a Roof Collapse, The Structural Engineer, Vol 64A, No 3, 1986.


1. Monograph on Planning and Design of Tall Buildings, Volume CL, Tall Building Criteria and Loading, American Society of Civil Engineers, 1980.

2. Civil Engineer's Handbook, Butterworths, London, 1974. 3. Bridge Aerodynamics Conference, Institute of Civil Engineers, Thomas Telford, London, 1981. 4. On Methods of Load Calculation, CIB Report No 9, Rotterdam, 1967. 5. BRE The Designer's Guide to Wind Loading of Building Structures

Part 1 Butterworths, 1985

Part 2, Butterworths, 1990.

Loaded Areas α

[kN/m2]

Category A

Category B

Category C

Category D

- general

- stairs

- balconies

- general

- stairs, balconies

- with fixed seats

- other

- general

2,0

3,0

4,0

3,0

4,0

4,0

5,0

5,0

Table 1 Imposed loads on floors in buildings

Snow load so [kN/m2]

Altitude [m]

Zone 0 200 400 600

1 0,40 0,49 0,70 0,95

2 0,80 0,98 1,40 1,89

3 1,20 1,47 2,09 2,84

4 1,60 1,97 2,79 3,78

5 2,00 2,46 2,49 4,73

Table 2 Snow loads for zones given in Figure 16

so = 0,412z

where:

A is the altitude of the site above mean sea level [m]

z is a constant, depending on the snow load zone.

Lecture 1B.4.1: Historical Development of Iron and Steel in Structures

OBJECTIVE/SCOPE

To appreciate how steel became the dominant structural material that it is today, it is essential to understand how it relates to cast iron and to wrought iron, both in its properties and in the way that all three materials evolved.

PREREQUISITES

None.

RELATED LECTURES

Lecture 1A.2: Steelmaking and Steel Products

SUMMARY

The properties of the three ferrous metals, cast iron, wrought iron, and steel, are described and the evolution of their production is summarized. The evolution of their structural use is also given and the prospects for further development introduced.

1. PROPERTIES OF THE THREE FERROUS METALS: CAST IRO N, WROUGHT IRON AND STEEL

Cast iron, as the name implies, is "cast" or shaped by pouring molten metal into a mould and letting it solidify; a wide variety of often very intricate forms is thus possible. It is very strong in compression, relatively weak in tension, much stiffer than timber, but brittle.

Wrought iron is strong both in tension and compression and ductile, thus making it a much safer material for beams than cast iron. Its main disadvantage is that, never reaching a fully molten state, it can only be shaped by rolling or forging, thus limiting its possible structural and decorative forms.

The properties of mild steel are similar to those of wrought iron but it is generally stronger and can be cast as well as rolled. However, it has a lower resistance to corrosion than wrought iron and is less malleable and thus not so suitable for working into elegant, flowing shapes.

These properties, in terms of strength and carbon content, are shown in Figure 1; the values shown should be considered as indicative rather than absolute limits. They do not include malleable or ductile cast irons which have strengths in tension considerably above those shown.

2. EVOLUTION OF FERROUS METALS

2.1 Blacksmith's Wrought Iron

Iron has been known and used for more than three thousand years, but it was not until the development of the blast furnace around 1500 AD that it could be produced in molten form. In China, molten iron goes back much earlier but this is not generally thought to have been known in the Western World until well after the independent invention of the blast furnace. There is slender evidence that the Romans knew how to produce cast iron but, if they did, the knowledge was certainly lost.

Before the blast furnace, iron was extracted from ore by chemical reduction in simple furnaces or hearths. Inevitably, the scale of the operation was small and the process quite laborious, the iron coming in a hard pasty form, far from liquid, which was then refined and shaped by hammering. Essentially, this was 'blacksmith's iron'.

2.2 Molten or Cast Iron

Although possible in the 16th Century, molten or cast iron was hard to produce on a large scale before the change from charcoal as a fuel to coke. With charcoal, the practical size of furnace was limited by the crushing of the fuel by the weight of the charge of the ore and thus the stifling of the blast. Abraham Darby I is generally credited with the mastery of coke smelting and, even though this was in 1709, coke smelting did not dominate the industry until about 1750 in Britain and considerably later in other parts of Europe.

2.3 Industrialised Wrought Iron

Large scale wrought iron, as opposed to blacksmith's iron, became possible mainly as a result of the developments culminating in Henry Cort's puddling furnace patented in 1793. In this furnace, the carbon in cast pig iron was burnt off in a reverbatory furnace while the impurities were drawn off by 'puddling'. As the process continued and the iron became purer, its melting point rose and the furnace charge became more viscous, eventually being removed in a stiff plastic form for rolling or forging. It was the enlarged

scale of the operation which was significant rather than any change in the actual material which was effectively the same as the blacksmith's variety.

The modernising of wrought iron depended not only on the puddling process, but the idea of grooved rollers which made possible the economic production of angle and tee sections, and later channels and joists. Here again, Henry Cort, who patented the grooved rollers in 1784, gets the credit although the due financial rewards eluded him.

2.4 Steel

Although steel-type iron had existed for many centuries, steel as used today dates from the 18th Century. It was produced either by cementation, a process by which bars of pure wrought iron absorbed carbon during prolonged heat treatment, or after about 1750 in molten form by Hunsman's crucible process. Cementation was largely confined to the cutlery and tool trades and has no real relevance to construction. Crucible steel continued to be made, although at a decreasing level of production, until after the Second World War; however it is uncertain how much of this was used structurally in construction works.

It is a common fallacy that the use of steel dates from Bessemer's converter of the mid 1850s; not only did Kelly in America get there first with an almost identical process, but the amount of steel already being produced was quite substantial. Some 60,000 tons of steel were produced each year around 1850 in Britain alone which is far from negligible, except perhaps when compared with an annual world production of 2,5 million tons of iron in the same period. Bessemer's steel was certainly cheaper and could be made in larger quantities, but its quality was uncertain. It was not until the perfection of the Siemens-Martin open-hearth process in the 1880s that steel moved in a big way into the construction and shipbuilding industries.

Today, very little truly structural cast iron is being used and no wrought iron is being made. Steel is wholly dominant. There are, however, some signs of a limited revival of cast iron, particularly in the new ductile form only available since the 1940s.

3. ACHIEVEMENTS WITH STRUCTURAL IRON & STEEL

In looking at the structural achievements with iron and steel in the last 250 years, it is convenient to class these in relation to the period, or age, when each of the three ferrous metals was dominant. Inevitably, these periods overlap and it is significant that in each case it took quite a long time - up to 50 years - before what was found to be possible became commercially widespread. The periods are broadly as follows:

Cast Iron Period 1780-1850 (Columns up to 1900)

Wrought Iron Period 1850-1900

Steel Period 1880 - Present Day

These dates are essentially based on Britain where the iron industry was more developed in the first half of the 19th Century than elsewhere. In France, there was no real cast iron period, while in America both cast iron and wrought iron were comparatively little used before the middle of the 19th Century, after which there was a positive explosion in their application. Steel on the other hand, became popular at roughly the same time throughout Europe and America. Figure 2 emphasises how short the overall period of structural use of iron and steel has been in relation to man's knowledge of iron.

4. THE PERIOD OF CAST IRON (1780-1850)

Given availability, new materials are introduced either for greater economy or to solve specific problems.

4.1 Cast Iron Arched Bridges

All the early cast iron bridges were arched forms in which cast iron merely replaced masonry, the advantages being greatly reduced weight and horizontal thrust, economy and speed of erection. The first iron bridge of any magnitude was the famous Coalbrookdale one completed in 1779 and spanning some 33 metres (Slide 1), a structure full of apparent illogicalities mixing carpenter's and mason's detailing but still standing proudly today. The construction of this bridge was followed by a whole succession of cast iron arch bridges in Britain, including Thomas Wilson's Wear Bridge of 1792-6 with wrought iron strapping to the cast voussoirs and a span of 72 metres (Slide 2) and Rennie's Southwark Bridge of 73 metre span completed in 1819. The climax, but by no means the last, cast iron bridge, was perhaps Telford's Mythe Bridge at Tewkesbury (1823-26) with a span of only 52 metres but great lightness and total structural logic (Slide 3).

Slide 1

Slide 2

Slide 3

In other parts of Europe, cast iron arch bridges were a rarity until well into the 19th Century, the number of schemes greatly exceeding the number built. Le Pont des Arts in Paris of 1801-3 by Cessart was, perhaps, the most famous, now, alas, replaced by a not wholly convincing welded lookalike. There were several early cast iron arch bridges in Russia.

4.2 Cast Iron in Buildings

With all buildings, fire was a recurring problem with timber structures. It was almost certainly the reason for one very early application of cast iron, the columns supporting the vast cooker hood and chimney of 1752 at the Monastery of Alcobaca in Portugal. In Britain, cast iron was used in the early 1770s in churches, partly for the cheap reproduction of Gothic ornament, but also for structural columns. In Russia architectural cast iron was used extensively throughout the 18th Century but it is not clear to what extent it was also used to support floors and roofs.

It is hard to see any trend arising from these early applications of iron to buildings. It was in the multi-storey textile mills in Britain in the 1790s that cast iron was first shown to have a major future in building structures. The disastrous fire at Albion Mill in 1791 was perhaps the biggest incentive for change. Bage and Strutt were the great pioneers. Between them, they developed totally incombustible interiors in cast iron and brick but with floor spans still of only about 2,5 to 3,0 metres in each direction, as had been the case with timber interiors. Later, this iron mill construction spread to warehouses with a gradual increase of spans.

While fire was the main reason for change in the mills, there was a growing desire in public buildings and large houses for long-span floors which did not sag or bounce. Timber had generally proved inadequate for spans above 6-7 metres. Between about 1810 and the early 1840s there was an increasing interest in cast iron floor beams, some with spans of 12 metres or more such as those in the British Museum of the early 1820s (Figure 3). Sometimes these castings were used as simple substitutes for the main timbers in essentially timber flooring, but in other cases brick jack arches, as in the mills of around 1800, or stone slabs were combined with long span cast iron beams to give rigidity, sound insulation and fire protection. Another form of 'fire proofing' consisted of wrought iron plates within the ceiling space arching between the cast iron beams. The climax of the development of cast iron flooring was reached in Barry's Palace of Westminster of the 1840s. Up to the mid 1840s, cast iron was seen as the wonder material everyone was looking for.

It is tantalising how little is known about who actually fixed the size and shape of the beams used by Nash, Barry and other architects of this period. Thomas Tredgold's book on cast iron of 1824 was undoubtedly influential but dangerously in error in some respects. In most cases, it is probable that proof-

loading of beams, which was widely used, provided the main safeguard against misconceptions and poor workmanship.

Apart from the mills and the long span floors, there was a whole range of new uses of cast iron between 1810 and 1840, sometimes on its own for complete structures as in Hungerford Market of 1836, or Bunning's highly decorated Coal Exchange of 1847-49. In Russia, there was also a considerable quantity of cast iron building construction in the first half of the 19th Century, as in the Alexandrinsky theatre of 1829-32 and the Dome of St Isaacs Cathedral (1837-41).

Towards the close of the 1840s, cast iron had lost much of its golden image and was being seen as an unreliable material, especially for beams. The progressive collapse of five storeys of Radcliffe's Mill in Oldham in 1844 and the failure of the Dee Bridge in 1847 were both highly damaging to its image.

4.3 Composite Cast and Wrought Iron in Building

Not all iron in the 'cast iron period' was cast. Some of it was composite cast and wrought iron and some simply wrought iron. There is little evidence of steel being used structurally in this period.

In Britain, cast iron was sometimes used in combination with timber as at New Tobacco Dock of 1811-14 or with wrought iron, as in the 1837 roof at Euston Station (Slide 4).

Slide 4

After 1840, the scale of iron construction and the proportion of wrought to cast iron in composite structures, increased substantially. The Palm House at Kew 1844-47, by Richard Turner and Decimus Burton, was a marked advance on earlier glasshouses and arguably incorporates the world's first rolled I sections. Wrought iron roofs of increasing span on cast iron columns proliferated both in the naval dockyards and for railway stations culminating in Turner's roof of 47 metres span at Lime Street, Liverpool (1849).

In France, some highly innovative wrought iron floors and roofs had been built before the Revolution, such as Victor Louis's 21 metre span roof of 1786 at the Palais Royal Theatre in Paris (Figure 4). In this roof, as in the case of the bridge at Coalbrookdale, the structural logic is not altogether clear. However, the flooring system of arched wrought iron flats devised by M. Ango in the 1780s (Figure 5) is clearly

understandable and derivatives of this system continued in use until they were largely replaced by a number of 'fire-proof' systems, still based on wrought iron, in the late 1840s. Cast iron impinged in France to quite an extent in the 1830s and after, notably in the great iron roof of 1837-38 at Chartres Cathedral and the Bibliotheque St Genevieve 1843-50, but it seems that wrought iron always retained its dominance.

Composite construction featured quite widely in Russia. In St Petersburg, a form of riveted plate girder was devised in 1838 for the repair of the Winter Palace after the fire of 1837. This development was just ten years before the independent development of riveted wrought iron beams in Britain.

4.4 Suspension Bridges

Some of the most creative work on the suspension bridge dates from the 'cast iron period' but is wholly related to wrought iron, although Tredgold did have the temerity to suggest cast iron support cables. In

most fields of construction, America clung to timber rather than iron in the first half of the 19th Century, but must be given credit for introducing the level deck suspension bridge, as patented by James Finley in 1808 with wrought iron chassis (Slide 5). Thereafter, there was a minor battle of principles on the form of cable. Britain favoured wrought iron chains with eye-bar links, as had Finley, while the French preferred wire cables, the difference being largely due to the states of the iron industries in the two countries.

Slide 5

By 1850, France had built several hundred suspension bridges, mainly due to the enterprise of the Seguin brothers, while Britain could claim scarcely more than a dozen. If the French had confined the wires to the sections of the cables above ground, all might have been well, but they did not. Corrosion became a major problem brought to a head by the collapse in 1850 of the Basse-Chaine suspension bridge with a death toll of 226. Thereafter, substantial remedial works followed and the building of suspension bridges all but stopped in France for many years. Nevertheless, based on French influence, wire cables did take over from eye bar chains in America and became virtually standard throughout the world.

5 THE WROUGHT IRON PERIOD (1850-1900)

5.1 Wrought Iron in Bridges

The wrought iron period was primarily the period of the riveted wrought iron beam which dates from the late 1840s, although by then wrought iron had established a fairly firm position in composite construction. Seen in the long term, wrought iron beams owe their birth, in part, to growing doubts both on the safety of cast iron in bending and in part to successful experience with iron ships. However, by far the biggest single contribution, not only to the development of riveted beams, but to the whole establishment of wrought iron as the dominant material of the period, was the design and construction of the Britannia and Conway tubular bridges, particularly the former.

The key figures here were Robert Stephenson, engineer to the Chester and Holyhead Railway; William Fairbairn, the practical man with experience of iron ships; and Eaton Hodgkinson, the theorist and experimenter.

Faced in 1845 with the then seemingly impossible task of taking trains over the Menai Straits, when shipping interests ruled out arches and suspension bridges as they had been shown to be inadequate for railway loads, they developed a new structural form, the box girder, and demonstrated it on a large enough scale for trains to run inside (Slide 6). However, it was not the bridges which mattered so much as

the understanding which resulted from the crash programme of research and testing which made them possible.

Slide 6

Between them, these three men dispelled the initial belief that wrought iron was weaker in compression than in tension, proved that a rectangular tube was stronger in bending than a circular or oval one, isolated the problem of plate buckling, and showed how to counteract this behaviour with cellular flanges and web stiffeners. Thus, these three men and their assistants established riveted wrought iron as a calculable material for beams of almost limitless size. Further, they demonstrated the benefits of continuity in beams, even for deadload (based on theoretical work from France) and proved that the strength of rivets depended on clamping as much as on dowel action. The extent of material and model testing for these bridges was prodigious.

The speed of the work was almost as remarkable as the result. The problem of crossing the Menai straits was posed early in 1845, the Conway Bridge was opened in December 1848 and the Britannia Bridge in March 1850. In both cases, work on the supporting masonry started in the spring of 1846 well before all the problems of the spanning structures had been solved. Other smaller wrought iron bridges of the same period, with cellular compression flanges were, it seems, all spin-offs from this basic development.

It is, perhaps, worth noting that concurrently with this major innovative work, Stephenson was responsible for a mass of other railway construction, including the six-span Newcastle High Level Bridge with cast iron tied arches of 1846-49 (Slide 7) and the ill-conceived Dee Bridge at Chester based on trussed

Slide 7

cast-iron beams, which collapsed disastrously in 1847 soon after it was opened. The pressure on the leaders of the engineering profession at this time are hard to imagine and it is no surprise that, sometimes, relationships became strained, as they did between Stephenson and Fairbairn.

The evolution of the plate girders of today from these beams with cellular compression flanges took place largely in the 1850s. Figure 6 shows some steps in this transformation.

The rationalisation of truss forms and their full structural evolution is another feature of the 1850s. Many of these forms derived from timber construction in America but given riveting and wrought iron the scope opened up enormously. The Britannia Bridge has been criticised for wasting material in comparison to an equivalent structure with open trussed sides, but this is unfair when one considers how little was known about true truss action in the mid 1840s. Figures 7a and 7b show typical intuitive and mathematically rational truss forms of this period. There were many variations on these forms.

Numerous wrought iron bridges of all forms and sizes followed in all countries. In Britain, I.K. Brunel's Saltash Bridge completed in 1859 and Thomas Bouch's fatal Tay Bridge opened in 1878, stand out for very different reasons. In France, Gustave Eiffel's great arches at Oporto and Garabit, of 1875-7 and 1880-84 respectively, are now world famous. In America, Charles Ellet's Wheeling Suspension Bridge of 1847-9, Roebling's Niagara Bridge completed 1855, and James Ead's St Louis Arch Bridge of 1867-1874 are all rightly famous, although one must add that the last of these is partly of steel.

5.2 Wrought Iron in Buildings

In buildings the scope for drama in the use of iron was generally more modest, the largest outlet being in flooring systems both in Britain and in other parts of Europe. It was almost certainly the development of these flooring systems in France in the late 1840s and early 1850s which provided the impetus for the commercial development of rolled joists, regardless of whether the first ones of all were rolled there or in Britain. The size of the joist sections gradually increased but until liquid steel took over, size was limited by the problems of handling large quantities of puddled iron.

Cast iron continued to be used extensively for columns well after 1850. In America there was a great vogue for cast iron facades which lasted for several decades. Bogardus and Badger were the two main suppliers. Internally, the structures vary, with iron, masonry and timber all represented.

Apart from these useful, but often unseen, applications of iron to traditional buildings, some spectacular iron build structures, mainly long span roofs, were built in all countries. Most commonly, but far from exclusively, they were over railway stations. They included the ribbed iron dome of the British Museum

Reading Room (1854-57), the 73 metre wrought iron arches at St Pancras Station (1868) and the dome of the Albert Hall (1867-71). These buildings were matched in France, for instance, by the Bibliotheque National (1868), Les Halles (1854-68) and the Bon Marche Department Store (1867-78); and in America by the dome of the Capitol in Washington (1856-64).

Throughout this period most buildings, particularly those of more than one storey, depended on masonry walls for stability, whether or not the floors and roof were of iron. The route to full structural framing in iron or steel is uncertain. It is often stated that the Home Insurance Building in Chicago of 1884-85 was the first fully framed tall building which formed part of a continuing development. Perhaps the earliest example of a stiff-jointed frame was Godfrey Greene's four-storey Boat Store at Sheerness of 1858-60. The Great Exhibition Building in London of 1851 and the Chocolat Menier Factory outside Paris of 1870-71 have also been claimed for this 'first', but they both had diagonal bracing and, anyway, had no apparently direct influence on the multi-storey steel construction of today.

6 THE STEEL PERIOD (1880-PRESENT DAY)

Steel is not only stronger than wrought iron, but being produced in a molten state made larger rolled or forged units practicable. However, it is not easy to identify which is which; for several decades, steelwork was fabricated by riveting in the same way as wrought iron and, when riveted, the two look almost exactly the same. The Forth Bridge in steel and the Eiffel Tower in wrought iron, were completed at almost exactly the same time (1889-90). Looking at them, who could tell the difference?

Figure 8 shows how steel took over in quantity from wrought iron in Britain. Figure 9 shows how the proportion of open-hearth steel increased until it had all but cornered the market by 1920. The biggest incentive for change to steel lay in the ship-building industry. Lloyds Register allowed steel plating of 4/5 the thickness of wrought iron and, by 1908, Lloyds was insisting that all steel for shipbuilding should be produced by the open-hearth process.

In bridges, the steel period was mainly one of increasing size and span. Here the initiative shifted away from Britain mainly to America where the need for major bridges, was greatest at this time. All the great suspension bridges up to 1945 (Golden Gate, George Washington, Transbay, etc.) were built of riveted steel with spun cables of high tensile steel wire.

In buildings the 'Skyscraper' came of age in steel, again with the initiative mainly in America. Long span roofs also took a leap in scale with steel both in France and America. First there were the great three-pin arch structures over the Philadelphia railway stations of 1893 (79 and 91 metre spans) followed by the Galerie des Machines for the 1889 Paris Exhibition of 111 metres span - over 50% up on St Pancras. These spans, in turn, have been dwarfed by the post-war domes over sports arenas. The span of the Louisiana Superdome of 1975 at 207 metres is more than 3½ times that of the Albert Hall.

The one big change in technique with steel was the introduction of welding, mainly from the 1930s, although possibly earlier. Today, the rivet is as dead as the production of wrought iron. Now welds and bolts dominate all construction in steel.

In all fields, new developments tend to follow new needs and this certainly seems to have been the case with bridges. Since the Second World War, most new thinking on suspension bridges, especially aerodynamic design and weight-saving, has been in Britain while Germany has led the field on the design of cable-stayed bridges.

7. PRESENT TECHNIQUES AND FUTURE PROSPECTS

One of the most noticeable moves in construction in the last ten years, in Britain certainly, but it seems elsewhere in Europe as well, has been towards a revival of structural steel for bridges and buildings. Fashions change in constructions, as in clothing, and so do needs and costs. It is, thus, interesting to look at some of the recent variants on normal structural steel and at rival materials to see how they have fared and to speculate on what may happen in the future.

Weathering steel (unpainted with stabilised corrosion) and exposed steelwork fire-proofed by water in hollow sections are both innovations of the 1960s but neither shows signs of wide adoption. On the other

hand, stainless steel, although in itself much more expensive than mild steel or even high tensile steel, is being found to be increasingly worthwhile when the cost of maintenance is considered.

Plastics have yet to make any significant impact except as a protective coating or for architectural trim.

Aluminium was once thought to be a dangerous rival to structural steel but, so far, it has made little impact in bridge or building structures. Reinforced concrete - still dependent on steel - has been a strong and growing competitor of fabricated steelwork since the 1890s, largely because of its in-built fire resistance, helped in the 1950s and 1960s by an architectural desire to 'expose the structure'. This trend is now being reversed and, since 1980, there has been a vigorous rebirth of structural steel. The increasing use of structural steel has been encouraged by the pursuit of 'fast-track' construction and the realisation that reinforced concrete is not a maintenance-free material. There has also been a swing in taste from visually expressed concrete to 'high tech' styling or to the complete wrapping of buildings in glass or masonry.

Future developments with structural steel in buildings are likely to be associated with fire protection. Thin intumescent coatings which froth up when heated and form a protective layer, are becoming still thinner - more like paint - but the need for such protection may be substantially reduced by the development of fire engineering. This development could lead to a new era of exposed steelwork with increasing attention to the shape and form of members and the appearance of joints. Castings of steel or ductile iron could well be in demand once more.


• The use of iron and steel in structures evolved through development in the production and properties of the three ferrous metals, cast iron, wrought iron and steel.

• Cast iron is formed into its final shape from molten metal a liquid which is poured into a mould and solidifies. Wrought iron never reaches a fully molten state and is shaped by rolling and forging. Mild steel can be cast as well as rolled but has a lower resistance to corrosion than wrought iron.

• Iron has been known and used for more than three thousand years but it is only in the last 250 years that new production methods have allowed the large scale use, first of cast iron, then wrought iron and finally steel. Cast iron was widely used in bridges and buildings in the period between 1750 - 1850.

• Wrought iron became popular during 1850 - 1900 allowing the construction of many novel bridges and building structures of increasing size and span.

• Steel came into increasing use from about 1880, and being stronger than wrought iron, has been used to build even larger structures. The introduction of welding of steel was a major innovation in connection techniques which facilitates the wider use of steel.

• For the future, stainless steel is being found to be increasingly attractive despite its greater cost. The development of fire engineering may lead to a new era of exposed steelwork together with a wider use of coatings of steel or ductile iron.

Lecture 1B.4.2: Historical Development of Steelwork Design

OBJECTIVE/SCOPE

To outline the developments in the design of iron and steel for structures.

PRE-REQUISITES


RELATED LECTURES

Lectures on the metallurgy of steel; a useful background to many other lectures, notably those dealing with the design of particular structural types.

SUMMARY

Structural theory as known today owes most of the intellectuals of France while in the late 18th Century and the early part of the 19th, Britain took the lead in practical design and application. 18th Century empiricism was replaced first by large-scale proof-loading and tentative calculation, followed after 1850 by component testing allied to elastic analysis with testing soon relegated to quality control. In the late 19th Century, the powerhouse of engineering thought shifted gradually to France, Germany and America. Elasticity and graphical analysis held sway for about 100 years until they were challenged by plastic theory and the computer, with automation replacing hand work in production and erection.

The developments in materials, theory and technique were all related but varied from country to country due to different needs, shortages and opportunities. This lecture outlines the developments in design methods for structural steelwork, illustrating this with a number of examples of iron and steel structures.

1. HISTORICAL DEVELOPMENT OF STEELWORK DESIGN: STAT E OF STRUCTURAL KNOWLEDGE IN THE 18TH CENTURY AND BEFORE

Up to the late 18th Century, structures were designed essentially on the basis of proportion. To some extent, this meant no more than deciding whether sizes looked right - that is, familiar - but in many, perhaps almost all periods, there were some rules or statements by authorities which were almost as firm as our codes of practice today. The difference is that they were not based on strength or stress but on shape and scale. Stress, in the sense that the word is used in engineering today, did not exist. The materials were essentially masonry and timber with a little iron.

With masonry the real problem has almost always been one of stability rather than crushing of the material and, until quite recently, stability was usually established visually. Early tie-bars of iron in masonry construction were, it seems, also sized by eye.

With timber in the 18th and early 19th Centuries, deflection was the main problem. If it was stiff enough, it must be strong enough. This may seem illogical to us today but with timber, which tends to indicate its distress by creaking, sagging and even splitting long before failure, stiffness was not a bad criterion for adequacy. Nevertheless, timber floors did sometimes collapse, perhaps most often due to ill-conceived joints.

Until the early 19th Century it is far from clear who fixed the sizes of timbers or the connections in trusses. Probably, it was the carpenters working on experience, observation and possibly copy books of details. In spite of growing knowledge of the strength and stiffness of different materials, this unscientific approach sufficed for the majority of construction until well into the 19th Century - at least in Britain, but perhaps less so in other parts of Europe.

2. STATE OF STRUCTURAL KNOWLEDGE IN BRITAIN IN THE EARLY 19TH CENTURY

In the early 19th Century, intuition gave way to calculation for all materials and theory took over to an ever increasing extent. However, the aim of this lecture is not to outline the development of structural theories for which most credit must go to the intellectuals of France, but to show how, in Britain particularly but also elsewhere, these theories were gradually incorporated in the work of ordinary engineering designers.

The fact that some of the theories were incorrect was of no importance provided that these were related to tests and that like was being compared with like. For instance, having established that for a rectangular beam the bending strength was proportional to:

(bd2) x (a constant depending on the material)

where b and d are breadth and depth of section, respectively, it did not matter whether you used Galileo's or Mariotte's incorrect theories of the 17th Century or Parent's elastically correct one of the 18th, provided that the constant was derived from bending tests and used in comparable circumstances for the assessment of the bending strength of other cross-sections. In 1803, Charles Bage developed a perfectly valid method of designing cast iron beams on the basis of tests and Galileo's bending theory.

Among the earliest mathematical design handbooks in Britain, if not actually the first, were Peter Barlow's book on timber, originally issued in 1817, and Thomas Tredgold's books on timber and cast iron, first issued in 1820 and 1822, respectively. Both Barlow and Tredgold made acknowledgements to earlier work by Girard and others on the Continent. It is worth looking quickly at the methods advocated in these books to get some idea of how at least a British engineer could have tackled the problems of fixing the sizes of structural members in the 1820s. The extent to which these handbooks were actually used is uncertain.

3. UNDERSTANDING OF TIMBER IN THE EARLY 19TH CENTUR Y

Much of the present practice with steel derived originally from timber which makes a good starting point.

In the simple case of direct tension, Barlow used the word 'cohesion' which is 'proportional to the number of fibres or to the area of section'. He tabulated 'cohesion on a square inch', as did Tredgold, both basing their values on their own experiments or those by Musschenbroek, Emerson, Rondolet and others. Thus, for direct force, the concept of stress was there in all but the name.

For timber, Barlow stated in relation to 'absolute strength' that 'practical men assert that not more than one fourth of this ought to be employed' but implied that so large a reduction was not necessary. Neither the effect of knots and other defects nor the concept of an overall factor of safety to cover all variables seemed to come into his thinking. Tredgold merely accepted a factor of safety of 4 on the ultimate strength of timber as disclosed by tests.

With timber, there was little need to consider beams of anything other than rectangular section. Barlow and Tredgold gave practical rules both for strength and deflection. For instance, for a rectangular beam of length L with a load of W, Tredgold's rule for strength amounted to:

W =

where the constant C allowed for the strength of the material, the loading conditions and different units for length and cross-section. There was no reference to bending moments or section moduli. All was direct, the tabulated values of C being derived from tests on small sections of comparable timber loaded in the same way.

It is notable that both Barlow and Tredgold devoted as much space to deflection as to strength, a clear follow-on from the time when sagging was the first and, perhaps the only, indication of inadequacy.

Tredgold suggested 1 in 480 as a reasonable limit for deflection in relation to span.

When considering floor joists, Tredgold's emphasis on deflection was particularly strong. He gave a rule, again controlled by a mysterious constant, which rightly relates the span, spacing and breadth of the joists to the cube (not the square) of their depth but, curiously, is independent of the load. He explained that the constant was based on scantlings 'found to be sufficiently strong' whereas 'it is difficult to calculate the weight that a floor has to support'. Thus, in this field anyway, the dominance of strength rather than proportion was not yet complete.

4. UNDERSTANDING OF CAST IRON IN THE EARLY 19TH CEN TURY

For cast iron, Tredgold, who certainly produced the first real calculator's guide to the material, moved closer to modern thinking than in his book on timber, but in some respects went very wrong, although pardonably so.

Again, he advocated a deflection limit of 1 in 480 for beams but also what we would call a safe working stress (f) of the frighteningly high value of 106 N/mm2 (6,8 tonf/in2). This value he considered to be the elastic limit in bending (based on tests on 25 x 25mm bars of cast iron). He also found the 'absolute strength of cast iron bars to resist a cross-strain' (modulus of rupture) of these small bars to be 280-400 N/mm2 and thus thought he had what amounted to a factor of safety of 2,6 to 3,8.

He then assumed, or so it seems because he said very little directly about it, that using the same working stress (f) in direct tension he would have a similar margin of safety as in bending. He assumed further and with more justification that using this stress (f) again in compression, the safety margin would be at least as high. Thus all one needed to do was to design to the elastic limit as a working stress and all would be well.

In the case of direct tension, Tredgold discounted the testing techniques which had given ultimate tensile strengths of around 110-120 N/mm2 and had no reason to know that later bending tests on larger beam castings were to show a modulus of rupture of as low as 110 N/mm2 for comparable iron. The last of these errors was specially understandable because the variation in the modulus of rupture with size of casting has still not been fully explained. Nevertheless, his thinking led to a potentially dangerous set of assumptions. He even suggested cast iron links at his universal working stress of 106 N/mm2 as more robust than wrought iron ones for suspension bridges.

It must not be implied that Tredgold got it all wrong. His method of calculating deflection appears to be generally correct. Further, with cast iron, there was a demand for cross-sections other than rectangular and Tredgold went into the properties of these sections at some length, getting the right answer with the symmetrical ones, but possibly not for quite the right reason, and going only slightly astray on the position of the neutral axis with T-sections and similar shapes.

On cast iron columns, as on timber ones, Tredgold's recommendations were basically sound. He was certainly aware of the problem of buckling and Timoshenko gives him credit for being the first to introduce a formula for calculating safe stresses for columns (see comparison in Figure 1). However, for ties he got into a tangle once more on the effect of length. He thought long ties to be stronger than short ones, visualising them as being subject to something like buckling in reverse which increased their strength with length.

The sad point about Tredgold's safe working stress, apart from his curious error on direct tension which had only a limited effect, is that if it had been applied to wrought iron it would have been almost universally sound. Also it would have been well ahead of any other practical guidance of the time, at least in Britain. The detailed thinking behind some of Tredgold's methods is not always easy to understand today, and it is doubtful whether many of his contemporary readers succeeded or even tried to follow this in detail. It is even more doubtful how many engineers in Britain read or understood the writings of men like Thomas Young or John Robinson or the works of the vast galaxy of theorists in other parts of Europe. Some certainly tried and the level of success would be hard to measure today.

Tredgold's book on cast iron was translated into French and German and ran into five editions, with the same errors perpetuated, the last being issued in 1860. However, from the 1830s onwards his practical advice was challenged by Eaton Hodgkinson's advocacy of his 'ideal section' for cast iron beams and his simple formula related to this.

Eaton Hodgkinson showed by direct loading tests that cast iron was about six times as strong in compression as in tension and proportioned his beam accordingly. His simple formula (Figure 2) has been repeated in engineering handbooks until well into this century.

All was derived from bending tests and would be equivalent to saying:

Ultimate resistance moment = N.D.A.t

where t is the ultimate tensile strength of cast iron. If N = say 0,9, the value of t derived from his formula would be 6,7 to 7,2 tonf/in2 which is a very plausible range. The significant point is that even Eaton Hodgkinson was not thinking in terms of stress but of a constant relating tests under one set of conditions to practical use in the same form. Eaton Hodgkinson also made extensive tests on cast iron columns and published the results with practical advice in 1840. This advice formed the basis for further recommendations for many decades.

5. UNDERSTANDING OF WROUGHT IRON IN THE EARLY 19TH CENTURY

Until towards the middle of the 19th Century, wrought iron was used almost exclusively in tension for such applications as chains, straps, tie rods and boiler plates.

The tensile strength of wrought iron was fairly well understood throughout Europe from early in the 19th Century, the mean value being about 400 N/mm2. Thus, even allowing for quite wide variations, its tensile strength could be relied upon to be about three or four times that of cast iron and with an incomparably greater ductility.

It was the behaviour of wrought iron in bending which eluded engineers until towards the middle of the 19th Century. There were, of course, the French wrought iron flooring units associated with Ango and St. Fart but these units were really tied arches.

Discounting the seemingly empirical wrought iron beams of 1839 (Figure 3) used in the Winter Palace at St Petersburgh which had no wider influence, the wrought iron beam dates from the mid 1840s when small rolled I beams were produced both in Britain and France. However, the really important breakthrough came from the research and testing for the Britannia and Conway tubular bridges. This work was a major achievement which, more than any other event, established the technique of building up structural sections of all sizes from rolled angles and plates by riveting. It made riveted wrought iron the premier structural material for almost 50 years. It also marked the climax of an era of component testing and proof-loading and heralded its end.

6. THE YEARS OF TESTING 1820-1850

Whatever may have been written about the strength of materials, engineers in this period tended to feel happier with tests than theory when facing new or uncertain conditions.

Proof-loading was widely applied to cast iron beams, in many cases all beams being individually tested. Records of important buildings indicate that the modulus of rupture under test often approached Tredgold's high figure of 106N/mm2. However unwise this figure may have been if the beams passed with a central point load, with the usual distributed loadings they must have had a factor of safety of 2 against the proof load.

Not only were full-size components such as beams and columns tested, but also small sections of different materials to establish their properties. Further, the development of new forms depended almost entirely on tests. Effectively the tubes for the Menai and Conway bridges were designed by experiment (Figure 4). Starting from the concept that wrought iron was just a less brittle form of cast iron, initial calculations were based on Eaton Hodgkinson's formula for cast iron beams. Tests then showed that unlike cast iron, wrought iron was apparently weaker in compression than in tension. Further tests proved that this was not a property of the material but due to plate buckling, a phenomenon not found in cast iron beams because of their heavy section. Other tests proved that for tubular beams, a rectangular shape was more efficient structurally than a circular or elliptical one, provided that its top and sides were stiff enough.

The tubes were designed for continuity over the intermediate supports even for self-weight (Figure 5) but it is not clear whether the continuity analysis in Edwin Clark's book of 1850 was used in the design or in

retrospect. Here again, modelling and testing probably paid a large part in the decision making. Irrespective of how the thinking may have developed, it led to the seemingly perfect form of a continuous tube with cellular top and bottom flanges, web stiffeners on its sides and trains running through the middle. At this stage, the form of web and flange stiffening seems to have been arrived at empirically. The tubular form of compression member gradually evolved into the simple I beam of today. Figure 6 shows some steps in this transition. It would, perhaps, be unfair to speculate on the amount of iron which might have been saved if the sides of the tubes had been open and triangulated. Such trusses could not then have been analysed, but nor, when work started, could riveted wrought iron box or I beams.

There is no space here to go into all the advances in understanding which accrued from the two year development programme for this seemingly impossible structure nor to try to disentangle the disputed contributions of Stephenson, Fairbairn and Eaton Hodgkinson. The more one looks at this stupendous achievement, the clearer it becomes that it was the testing which came first and showed 'how' and the

theory which followed up and explained 'why'. Engineers in Britain throughout the 19th Century were frightened of mathematics.

It is notable that in this same book with the analysis of continuity, Edwin Clark still felt constrained to say of 'transverse strain':

"The complete theory of a beam, in the present state of mechanical science, is involved in difficulties. The comparative amount of strain at the centre of the beam where the strain is greatest, or at any other section, is easily achieved but the exact nature of the resistance of any given material almost defies mathematical investigation".

Because of the magnitude of the achievement, we may be overestimating the understanding of those responsible. Certainly the dispute over the Torksey Bridge in 1850 showed that continuity was not widely understood.

7. TERMINOLOGY: STRAIN, STRESS, COHESION, ETC.

This may be the point where a short diversion on terminology is appropriate. In the first half of the 19th Century the word 'stress' virtually did not exist in engineering. What is referred to as stress today was called strain or sometimes, if tensile, cohesion, but 'strain' also seems to have been used to denote a force (e.g. a strain of 10 tons). There was some uncertainty in the use of these terms.

The relationship which really meant something was the proportional one between member size and load. If, in Tredgold's words, "the strain in lbs. a square inch which any material would bear was x then four square inches would bear 4x". That was alright for direct tension and compression but with bending, the explanations are less clear.

According to Timoshenko the concept of 'stress on an infinitesimal plane' was due to Augustin Cauchy and published in 1822. Cauchy also developed the valuable concept of principal stress but again, according to Timoshenko, it was St. Venant who first defined stress in its final form in 1845. Both Todhunter and Pearson, Timoshenko and others give W.J.M. Rankine the credit for being the first to provide rigorous definitions of stress, strain, working stress, proof strength, factor of safety and other phrases which are now commonplace in engineering.

8. STRUCTURAL DESIGN BETWEEN 1850 AND 1900

While there is a danger of over-elevating the Menai Bridge designers today, there is an even greater risk of assuming that their new-found understanding was immediately absorbed by all other engineers. It was not, but there was a very great change in attitude mainly in the years between 1850 and 1870. This was the period when ordinary engineers learnt to calculate the sufficiency of most simple structural forms, beams in particular, and to believe in their calculations - even for major structures - without testing.

1850-1870 was also the period when it became possible to analyse the forces in trusses with certainty. Several researchers contributed to the understanding of the forces in complex but determinate trusses. Practical textbooks were published in different countries and translated into other languages, all telling roughly the same story. Rankine's "Manual of Civil Engineering" (1859) was very widely read and frequently reprinted. W.C. Unwin's "Wrought Iron Bridges and Roofs" of 1869 showed how graphical statics now dominated truss analysis (Figure 7). Unwin and others also showed how to build up flanges and cover plates to match the bending moments (Figure 8). Another interesting practical textbook is that written by Professor August Ritter of Aix-La-Chapelle Polytechnic and published in 1862. This book

gives complete analyses of several notable British structures of wrought iron and was considered worth translating into English in 1878. Many of the methods of the 1850s and 1860s, although perfectly practicable, proved tedious until R.H. Bow introduced his famous notation in 1873. This was exactly the sort of systematic and almost foolproof graphical method to appeal to engineers. It has retained its popularity through many generations and has been superseded only recently for speed by the computer.

In spite of growing confidence, load testing took some time to die. Large scale tests were still being used around 1850-60 although possibly as much to satisfy clients as to reassure designers. In the late 1840s three of the crescent trusses of 47m span for the first Lime Street Station in Liverpool were erected as a unit in Turner's works in Dublin and tested first for a uniform load of almost 2kN/m2 and then for eccentric loading. These trusses have a record span and the need for assurance was understandable.

The proving of the 65m trusses for New Street Station in Birmingham (another record span completed in 1854) was even more elaborate, as show in Figure 9. Apart from testing the performance of a complete section of the roof, each tie member was proved to 139 N/mm2 before incorporation.

After about 1860, confidence in wrought iron had grown enough for testing even of major building structures to be played down, although bridge testing continued.

Provision was made for testing in the contract for St Pancras Station (completed 1868) but it was never used. The Albert Hall roof (1867-71) was erected in Fairbairn's works in Manchester to make sure it fitted together but was not load-tested. These are just examples. One could cite others to illustrate the change from intuition and physical verification to the calculation of sizes with confidence.

One reason for this change was, of course, the displacement of cast by wrought iron. Wrought iron was now recognised as a reliable material and, with rivets of definable strength it could be built up into structures virtually limitless in scale in spite of restrictions on the sizes of plate and angle which could be rolled. Further, and most important of all, by 1850 or soon after, it had become a calculable material, not just for ties and struts but also for beams.

While it was mainly the triumvirate of Stephenson, Fairbairn and Hodgkinson who established the riveted wrought iron beam, it was the 'elasticians' of the mid-century like Rankine who translated this knowledge into practical advice and showed engineers how to design with it.

With increased understanding of structural behaviour, there was a swing at this time from intuitive feelings that strength and stiffness could be increased by redundancy to simplification of forms so that they would be more amenable to precise calculation and thus to more economical sizing.

The reality of the known behaviour of wrought iron was limited to the range of stress within which the theorists were thinking. With a working stress generally not exceeding 77 N/mm2 (the Board of Trade figure in Britain) there is no doubt that wrought iron behaved elastically and that the theory of elasticity, which became the gospel for engineers in the third-quarter of the 19th Century, was wholly relevant.

Hooke's law held. Young's modulus was a constant. There was no need to think about factors of safety. You had a working stress to control your design, even though you might still have been calling it a strain, and you had every reason to feel confident.

Stress, as we understand it, had not only been born but, by now, was the controlling factor in almost all structural design, at least with iron, and iron was becoming increasingly dominant where a high level of performance was needed. Elastic theory, graphical analysis and definite rivet strengths were all that the designer required for full confidence. Around 1850, Britain had such confidence and was still leading the field in iron construction, although much was being done in parallel elsewhere, in particular in France, Germany and America. As the century progressed, the initiative moved from Britain with engineers like Moisant (Chocolat Menier Factory) and Eiffel and his colleagues catching much of the limelight.

The commercial transition from wrought iron to steel roughly between 1880 and 1900, permitted higher working stresses (generally 93 N/mm2 instead of 77N/mm2) and the use of larger rolled sections. Initially, it had virtually no effect on design and detailing.

Cast iron columns continued to be used widely until about 1890-1900 but were then superseded first by wrought iron but mainly by steel. Further theoretical work on buckling went in parallel with more advanced formulae for safe loads. It seems that amongst practising engineers the question of buckling of struts and of thin plates remained the least well understood aspect of structural design throughout the 19th Century.

It is not the intention of this lecture to chart the development of theoretical knowledge but rather to show how this related to the ordinary engineer in the design office. To follow the understanding of bending, shear and instability in more detail, the works referred to in the list of Additional Reading should be consulted.

9. POSTSCRIPT ON THE 20TH CENTURY

In the early part of the present century, the greatest advances both in theoretical understanding of structures and in practice were associated with the airship and aircraft industries. For bridges, buildings and other 'heavy' structures the changes were mostly associated directly or indirectly with welding.

The general introduction of welding in the 1930s (with Britain lagging behind other parts of Europe and America) radically altered techniques of fabrication and introduced the possibility of joints as stiff as the members they connected. This development in turn had its effect on design with more emphasis on 'portal framing' for buildings and stability through stiff joints rather than diagonal bracing.

The big change in design thinking came with plasticity in the late 1930s although ultimate-load thinking with the concept of the plastic hinge has taken some time to replace elastic theory. In fact, it has not wholly done so yet. Safe stresses are still quite dominant after a reign of nearly 150 years, but their use is declining.

In the future, engineers are likely to be able to achieve far greater efficiency by considering 'whole structure' behaviour including the effects of cladding and partitions especially for stiffness. This approach only becomes practicable with computers but offers attractive possibilities for the years to come. The disadvantage could be a reduction in adaptability. Also the understanding of designers needs to keep pace with the growing sophistication of the design aids at their disposal.


• Up to the late 18th Century, structures were designed essentially on the basis of proportion. • Intuition gave way to calculation for all materials and theory took over to an increasing extent in

the 19th century. • Much of the present practice in steel design derived originally from timber in the 19th century. At

that time the understanding of cast iron and wrought iron grew largely on the basis of component testing and proof loading. Rigorous definitions of stress, strain, working stress, proof loading and factor of safety appeared in the mid 19th century and gradually ordinary engineers learnt to calculate simple structural forms on the basis of assumed elastic behaviour and believe in the calculations without testing.

• In the 20th century, the greatest advances in the theoretical understanding of structures were associated with the airship and aircraft industries.

• The introduction of welding in he 1930s and the development of the theory of plasticity led to major changes in design thinking.

• For the future, the wider use of computers offers the possibility of achieving greater efficiencies in structures by considering 'whole structure' behaviour including the effects of cladding and partitions.


I Those who wish to delve deeply into the way in which structural theory as we know it today first emerged in the late 18th and early 19th Centuries, would do well to go straight to the classic authors: Coulomb, Bernouli, Euler, Navier and others.

For a more general view of structural theory and how it developed, the following books are recommended:

1. Timoshenko S P. "History of the Strength of Materials", McGraw-Hill, New York, 1953.

2. Todhunter I & Pearson K. "A History of the Theory of Elasticity and of the Strength of Materials from Galileo to the Present Time", Cambridge University Press; 3 volumes 1886-93.

3. Charlton T M. "A History of the Theory of Structures in the Nineteenth Century", Cambridge University Press 1982.

4. Mazzolani F. "Theory and Design of Steel Structures" Chapman & Hall, London.

5. Heyman J. "Coulomb's Memoir on Statics: an essay in the history of civil engineering", Cambridge University Press 1972.

II For a guide to practice with iron and later steel, there were many guides and text books published, especially after 1850.

Taken as a sequence, the following books give some idea of how this advice developed:

1. Tredgold T. "Elementary Principles of Carpentry", London: Taylor 1820.

The major British work on the structural use of timber, first published in 1820 and being reprinted as late as the 1940s. There are some details on the use of iron with timber, particularly for the lengthening and strengthening of timber beams.

2. Tredgold T. "Practical essay on the strength of cast iron and other metals", London: Taylor 1822.

Also several later editions.

3. Barlow P. "A Treatise of the Strength of Timber, Cast Iron, Malleable Iron & Other Materials", London: J Weale, 1837.

The 1837 and later editions were extensively revised and added to to take account of developments in the science of the strength of materials in the railway age.

4. Unwin W C. "Wrought Iron Bridges & Roofs", 1869.

Originally lectures to the Royal Engineer Establishment, Chatham.

5. Rankine W J M. "A Manual of Civil Engineering", London 1859, and later editions.

Rankine's manuals mark the turning point in Britain, of engineering as a science founded on theory as against an art founded on practical experience and observation. They summarise and extend earlier theoretical texts, notably on theory of structures and strength of materials, and remained standard works throughout the 19th Century.

6. Warren W H. "Engineering Construction in Iron, Steel & Timber", Longmans, London 1894.

Lecture 1B.4.3: Historical Development of Iron and Steel in Buildings

OBJECTIVE/SCOPE

To review developments in steel building construction, demonstrating how improvements in material and understanding have enabled greater achievements in terms of height, clear spans and building efficiency.

PREREQUISITES

None.

RELATED LECTURES



Lecture 1B.4.4: Historical Development of Iron and Steel in Bridges

SUMMARY

Iron was originally used for the principal components in building structures in order to achieve fire resistant construction. Initial forms followed traditional patterns, but gradually the characteristics of iron, and subsequently steel, were more fully utilised. Various building categories are considered - mill buildings, long span roofs, and multi-storey buildings. Significant technical innovations and design approaches are highlighted.

1. INTRODUCTION

Although the history of iron and steel dates back several hundred years, their use in the main components of building structures is relatively recent. The Industrial Revolution provided both the means and the need. Coke smelting and steam power enabled greatly increased production of iron, and the industrial mill buildings were foremost in the structural use of the material to replace timber. Inevitably, the adoption of a new material is spasmodic, and at times may even become unfashionable. Wrought iron, for instance, never totally replaced cast iron, any more than cast iron replaced timber. Any historical review will, therefore, include discontinuities rather than be a smooth sequential development. To simplify this review, the history is, therefore, subdivided by building type - mills and industrial buildings, long span roofs such as conservatories, railway stations and exhibition halls, and multi-storey frames. The development of new design forms to take advantage of improvements in material characteristics is traced for each type.

2. EARLY STRUCTURAL USES OF IRON IN BUILDINGS

Steel and before that iron, have been used in building construction for a very long time. The first uses were as secondary components - connectors, shoes and straps, mainly in combination with timber as the principal structural material. As early as the 6th Century, iron tie bars were incorporated in the main arcades of the Haghia Sophia in Istanbul. Domes often relied on tie bars to reinforce their base, such as in Jacques Germain Soufflot's portico of the Pantheon in Paris (1770-72). However, the most prominent early application of the material was in the decorative use of wrought iron, for instance, in balustrades and

gates. An outstanding example is to be found in the White Gates at Leeswood in Clwyd, Wales (1726) (Slide 8). Thomas Rickman combined the structural utility of cast iron columns with delicate ornament in the gallery fronts and ceilings to the nave and aisles of St George's Church, Everton, UK (1812-14) (Slide 9). In France, the architect Henri Labrouste designed two notable libraries. The Bibliotheque Sainte Geneviève (1843-50) (Slide 10) utilises cast iron for columns and arches to support both roof and floor, whilst at the Bibliotheque Nationale (1858-68) (Slide 11), the same decorative use is made of cast iron, but this time in combination with wrought iron.

Slide 8

Slide 9

Slide 10

Slide 11

These and some other early examples of public buildings which used iron exposed the structure in the interior but gave no sign of it from outside. J.B. Bunning's Coal Exchange in the City of London (1849) incorporated an iron framed galleried atrium behind two palazzo office blocks, while the Bibliotheque Sainte Geneviève had a scholarly Renaissance stone facade. Dean and Woodward used iron and glass extensively for their Oxford Museum (1860) (Slide 12) creating a dramatic interior.

Slide 12

3. INDUSTRIAL BUILDINGS AND MILLS

The introduction of iron components as principal structural elements is a relatively recent development, inspired by the desire for fire resistant construction. Earlier timber framed construction was always vulnerable to fire, particularly in textile mills where cotton fibres were processed in an oily, candle-lit atmosphere. By the end of the 18th Century iron was beginning to replace timber for the main structure. Initially, this was for the columns only, the first examples being a cotton mill in Derby, UK and a warehouse in Milford, UK (1792-93). The designer William Strutt used brick jack arches in place of the traditional timber floor. The jack arches sprang from iron plated timber beams with a plastered soffit to provide increased fire resistance. The beams were supported externally on the masonry walls and internally on cast iron columns.

The next logical progression was to use iron instead of timber for the beams. The first example of such a building frame was Charles Bage's Flax Mill at Shrewsbury, built in 1796 (Slide 13). The external masonry is loadbearing, but internally slender cast iron columns support cast iron lattice girders enclosed within brick arch floors. The building still stands today, having been used most recently as a maltings. The beams were cast in two sections, bolted together, with a skewback base, designed to carry brick arches. Their profile, which was concealed by the brickwork, rises at mid-span.

Slide 13

The combination of an external loadbearing masonry envelope and an internal iron frame became a common form in Britain, particularly for industrial buildings, such as the Albert Dock buildings in Liverpool (Slide 14). These buildings were constructed in 1845 and have recently been refurbished to provide office and residential accommodation. This period of structural design using iron was characterised more by evolution of form than by revolutionary new systems. Beam cross-sections saw the development of first the inverted T section (the bottom flange carrying the arch) and later the I section. Column sections also altered. Cruciform sections were superseded by circular hollow sections which could also accommodate steam heating or rainwater flow.

Slide 14

In 1856, Gardener's store (Slide 15) - an elegant furniture warehouse - was erected in Jamaica Street, Glasgow. This building used a cast iron frame system patented by a local ironfounder, Robert McConnel, for the facade, but the flooring system was based on a timber structure. The framing system allowed a rich expression of the fenestration, and was similar in principle to those first used in St Louis, USA.

Slide 15

The first building with a true rigid iron frame, making no structural use of loadbearing masonry, was Greene's Boat House completed in 1858 (Slide 16) at the naval dockyard, Sheerness, UK. This building was a four storey, three bay frame 64m by 41m by 16m high. The primary beams are of riveted wrought iron and span 9m. The secondary beams are cast iron and span 4m. Corner columns are hollow cast iron and are used as down pipes, whilst others are of H-section. The frame not only carried the full vertical loads but also provided the lateral stability.

Slide 16

In France, the first fully framed building was the Menier Chocolate Factory (Slide 17) at Noisiel-sur-Marne, completed in 1872. The most distinctive feature of this building, which is constructed over the River Marne which powered its machinery, is the diagonal bracing which is so elegantly (Slide 18) expressed on the exterior. This bracing provides the necessary lateral rigidity to the slender wrought iron skeleton, the decorative brick infill walls serving no structural purpose.

Slide 17

Slide 18

In Germany, an octagonal steel frame was used by Bruno Taut to support a gold coloured sphere in his design for the pavilion at the Leipzig Fair (1913) and Peter Behrens designed a steel three pin arch for the AEG turbine hall in Berlin (1909) (Slide 19).

Slide 19

The introduction of bracing systems freed the structure from its dependence on masonry walls for stability, and other materials began to be employed. Corrugated iron, the ancestor of today's profiled steel sheet (Slide 20) was patented in 1829. Forming iron into thin sheets with undulations to give stiffness was the idea of Henry Robinson Palmer who worked for the London Dock and Harbour Company. The corrugated sheets were manufactured by Richard Walker and were used on warehouse and storage buildings at the docks.

Slide 20

The combination of steel frame and lightweight cladding has continued to be a popular solution for industrial buildings. Many of the structural forms have been developed to create longer roof spans, and here the historical development merges with that of other building types.

4. LONG SPAN ROOFS

The developments in iron bridge construction were paralleled by those in long span roof forms. In 1786, Victor Louis designed a tied arch roof using wrought iron to span 21m over the Theatre Francais. He introduced many sophisticated features such as shaping fabricated elements to provide greatest resistance to bending and buckling and achieving a form which was both elegant and daring: qualities which characterised French iron structures for more than a Century afterwards.

Many of the early clear span iron structures borrowed ideas and principles from contemporary masonry and timber construction, such as the stone arch on which many cast iron bridges were based. Often timber structures destroyed by fire were replaced by iron structures of a similar form. Examples include the cupola of the Granary in Paris (destroyed by fire in 1802 and replaced in 1811) and the roof of Chartres Cathedral (1836) (Slide 21). Here cast iron was used by Emile Martin for the curved frames of the arching roof, but the tie rods at the springing were wrought iron. The roof spans 14,2m with a clear height of more than 10m from vaulting to apex.

Slide 21

In the first half of the nineteenth Century many innovative iron structures were built in France where technical, educational and scientific understanding were most advanced. Wrought iron was used for other long span roof structures in France, such as La Bourse (1823) (Slide 22). It is interesting to note that in Britain cast iron remained the favoured material for buildings constructed during the same period - for instance, the floors of Buckingham Palace and the floors and roof of the British Museum.

Slide 22

In both Britain and France, iron and glass were married in the construction of numerous glasshouses and conservatories, the slender glazing bars making iron an ideal material. Early examples include a palm house at Bicton, Devon (1816) (Slide 23) which uses a wrought iron glazing bar system devised by Loudon. This system established a pattern for glasshouse construction, and a later example is Turner and Burton's Palm House at Kew (c. 1847) (Slide 24) which uses curved ironwork throughout. Both of these examples have recently been restored. The latter is 110m long with a maximum clear span of 15,2m and raised to 19m at its centre. The structure of the main ribs is of curved wrought iron beams, as used in the construction of ships decks. The purlins, also of wrought iron, consist of a tensioned rod running within a pipe between ribs. The decision to substitute wrought iron for cast iron substantially reduced the weight of the structure and allowed greater light penetration into the building - a very important consideration in glasshouse construction.

Slide 23

Slide 24

Similar forms were used to build very long span roofs over railway termini. The roof at Euston station (1835-39) consisting of two 13m spans supported on slender cast iron columns, is believed to be the first example of all iron roof truss construction. The designer, Charles Fox, working under Robert Stephenson, used rolled iron T sections for the rafters and the compression members and rolled bar for the tension members. The connections were made by forging and drilling ends to the bars for bolting, with wedges used for adjustment. However, an accident at this station in which a derailed train demolished an internal column causing a partial roof collapse, led to the need for clear spans. Notable examples include Turner's Liverpool Lime Street, spanning 47m (1849) and Barlow's St.Pancras spanning 73m (1868) (Slide 25).

Slide 25

At Liverpool Lime Street the structure took the form of arched trusses, sliding joints at the supports preventing lateral thrusts being transferred to the supports and thus avoiding arching action. The construction of the roof was completed in 10 months. In contrast, St. Pancras uses a trussed arch with the outward thrusts at the springing contained by ties located below platform level. It is interesting to note that many of the designs for these long span roof structures were regarded as so innovative that the railway companies demanded full scale tests to demonstrate their integrity.

In France, Camille Polonceau developed a simple trussed rafter system using iron, sometimes in combination with timber. This system was widely used in a variety of building types, including the roofs over the Paris-Versailles Railway (1837). These trusses had timber principals, cast iron struts and wrought iron ties.

Paxton's Crystal Palace (1851) (Slide 26) was another remarkable structure built during this period. His design for the exhibition hall was a rectangular building 564m long by 22m wide and rising to a maximum height of 32m. It consisted of a framework of cast iron columns with cast and wrought iron trusses, connected using wrought iron and wood keys. However, much of the credit for this structure must go to the ironwork contractors Fox Henderson & Co. They were responsible for the structural analysis, working drawings and construction, bringing their experience on bridges, dockyard roofs and prefabricated buildings to enable completion of the building within a period of four months. Other major buildings by them include the trainsheds at Paddington (1851-4) and Birmingham New Street (1854).

Slide 26

In France, one of the most spectacular exhibition halls, the Galerie des Machines (Slide 27) was built for the 1889 Paris exhibition. It was the architect Dutert whose idea it was to enclose the 420m long, 110m wide hall with a single span. In conjunction with engineers Contamin, Pierron and Charton he developed the three-pinned, trussed steel portal frame, rising at its apex to a height of 43m. Like the Eiffel Tower (Slide 28), it was constructed from many small sections and plates riveted together in truss-like form. The purlins, too, were of lattice construction. The scale of the detail was enormous.

Slide 27

Slide 28

In America, too, the iron truss gradually gained favour, an early example being the Library of Congress in the Capitol Building, Washington (1854). However, it was the emergence of the mass production industries in the 1920s and their highly developed factory layouts which provided the opportunities for new structural forms, pioneered by Albert Kahn. The need for production flexibility dictated wide span industrial buildings. Deep lattice truss construction had been used for some time in bridge design and Kahn adopted this for many of his buildings. Natural lighting was provided in the production areas by adopting a monitor roof form. This improved lighting compared with north light roof forms, but avoided excessive heat gain.

Examples of this form of construction include the press shop for Chrysler at Detroit (1936) and the Assembly Building for the Glenn Martin Aircraft Company at Baltimore (1937). Trusses 9m deep spanned over 90m to give a column-free floor area of 150m by 100m. The monitor roof light was achieved by bridging alternately between the top and bottom chords of these trusses.

As spans became longer, so lateral stability of the trusses became more critical. This was countered by using box or triangular cross-section trusses. The trend towards longer spans led to the development of space frame construction which allowed advantage to be taken of the ability of such structures to span in two directions. In fact, the development of this has its origins in the work of Alexander Graham Bell at the beginning of the 20th Century. However, the first system widely available commercially, the MERO system, was not introduced until the 1940s. This structural form has proved a popular method for roofing long spans very efficiently, and other commercial systems have developed and continue to be used up to the present.

5. MULTI-STOREY BUILDING FRAMES

Just as iron was becoming more popular as a structural material for mill buildings and long span roofs, so too was it being increasingly used for multi-storey building construction. It was in North America and, in particular, Chicago that most development took place. Two important influences were the need to build higher to overcome the chronic overcrowding of cities of the period, and the terrible fire of 1871 which completely devastated the commercial quarter of Chicago. Another vital element in the development of high-rise construction was the introduction of the passenger lift by Elisha Otis in 1853.

Just as with industrial building development, changes in the form of construction took place in several steps. By the 1860s cast iron columns and wrought iron girders were commonly used to support brick arch floors, but with external loadbearing masonry still carrying a proportion of the vertical loads and providing lateral stability. William le Baron Jenney's First Leiter building (Slide 29), completed in 1879 in Chicago, for instance, is basically a hybrid with timber secondary beams, wrought iron primary beams, cast iron columns (internal) and masonry piers on the perimeter.

Slide 29

Before their general use for commercial buildings, tall iron frame structures began to appear towards the end of the 19th Century. Perhaps the most famous of these is the Eiffel Tower which remains as one of the most potent symbols of iron construction. Built as a temporary monument to crown the 1889 Paris Exposition, at 300m it was the highest structure of its time (although other similar towers had been proposed in cast iron as early as 1833). The design of the tower was, in fact, developed initially by Koechlin and Nougier, two engineers working in Eiffel's office. An architect, Sauvestre, also working for Eiffel, made important modifications including joining the first level and the four main legs with monumental arches. Eiffel, however, assumed responsibility for its construction.

Other notable structures of this type include the Latting Observatory Tower (1853) and Statue of Liberty (1886), both in New York.

It was not until about 1880 in the USA that the full potential of iron and steel frames was realised and they became standard for high buildings. The advantages of a frame structure can be seen by comparing the loadbearing masonry Monadnock Building, Chicago (1885) with the second Monadnock Building completed in 1891 using a steel frame. The walls of the earlier building measure 4,5m thick at their base. However, as late as 1890 loadbearing masonry was used for the Pullitzer Building, New York with walls 2,7m thick.

5.1 Floor Construction

It was recognised that substituting iron or steel for timber was not the complete answer to providing fire safety since unprotected iron beams would lose their strength at high temperatures and cast iron columns could fail when suddenly cooled by water from fire hoses. Some form of additional fire protection was, therefore, necessary. This requirement was clearly demonstrated by a plaster encased building structure which survived the Chicago fire.

The jack arch floor construction methods used earlier for mill buildings were largely unsuitable for resisting fire, partly because of their weight and partly because the lower flange of the iron beam would be exposed in the event of a fire. Terracotta flooring, in which hollow blocks of terracotta formed 'flat arches' to span between the lower flanges of the beams, overcame both of these problems. An early example of this form of floor construction is the 7-storey Tribune Building in New York (1869) which was also one of the first buildings to incorporate a passenger lift. Various systems based on this principle were developed. The blocks were arranged to project below the lower flange of the beam which was afforded fire protection either by projecting flanges of terracotta, or by cover slips of terracotta supported on small nibs. Floor finishes were either terracotta floor tiles or concrete, and their weight was about half that of the brick and concrete arch floors. This meant a significant reduction in the self weight of the structure and hence the load to be carried by the walls, columns and foundations, which was particularly important in Chicago with its poor subsoil conditions.

Other floor systems were developed using expanded metal as permanent shuttering but these needed separate ceilings. In 1846 the first iron beam was rolled in France, with the subsequent development of floor systems such as Système Vaux and Système Thuasne. These consisted of wrought iron beams at about 600-900mm centres connected by iron rods with a thick (70mm) plaster ceiling encasing the lower part of the beam. In Britain 'filler joist' floors, comprising closely spaced joists with concrete cast between, became common during the early part of the 20th Century. In many respects these floors can be seen as the precursor to the composite and reinforced concrete floor slab systems in current use.

5.2 Beams and Columns

The iron beams supporting the floors were initially formed as truss-like girders by riveting small cast or wrought iron elements. These girders were relatively deep and the planning was generally arranged so that they could be incorporated within partition walls. It was not until much later that rolling of wide flanged beams became possible, allowing shallower construction depth and hence greater planning freedom.

Cast iron columns remained popular for some time. It was not until the recognition of the need for bending strength within columns to deal with eccentric loads, that wrought iron, and subsequently steel, really took over. Like beams, the columns were initially formed by riveting a number of small sections to form a cross-section with similar bending strengths about both axes.

5.3 Frame Construction

The first move towards a fully framed form of construction was the introduction of columns within (or in front of) the external walls so that the masonry carried only its self weight and none of the floor loads. Only when the frame carried not only the floor loads but also the external wall was the height of construction no longer limited by the ability of the wall to carry its own weight. This arrangement also solved the problem of the differential thermal expansion of masonry and iron.

Jenney's 10-storey Home Insurance Building, Chicago (1885) is considered to be the first fully framed building to adopt this form of construction and as such was the first skeletal skyscraper. Cast iron columns support wrought iron beams for the lower floors and Bessemer steel beams above the sixth floor. The frame was fire protected throughout by masonry and fire clay tiles. The external walls were carried on angles attached to the spandrel beams, although this detail was not revealed until the demolition of the building in 1931.

Another early example was the 11-storey Tower Building in New York designed by Bradford Lee Guilbert in 1887 for a very narrow site. Loadbearing masonry walls would have been so thick at their base that no useable space would have been left.

5.4 Wind Braced Structures

Although these developments led to structural framing systems designed to carry the full vertical load including the self weight of the external walls, the structure was still dependent on the walls for lateral stability. The cross bracing used in the exterior of the chocolate factory at Noisiel-sur-Marne was generally regarded as inappropriate for commercial buildings, and the stiffness of the connections utilised in the Crystal Palace to provide stability was recognised as being inadequate for the more onerous demands of high-rise buildings.

The first Monadnock Building, although of loadbearing masonry construction, used a combination of portal frame bracing and masonry cross walls. Many other buildings used a mixture of methods.

Jenney's 16-storey Manhattan Building, Chicago (1890) was the first with a wind braced frame. This frame consisted of a combination of portal bracing and diagonal wrought iron rods tightened with turnbuckles. This building also provides an interesting commentary on the relative material costs at the time. Steel was used only for the major beams because of its high cost, with wrought iron for secondary beams and cast iron for columns.

Burnham and Root's 22-storey Masonic Temple (1892) was braced with diagonal wrought iron rods placed in the transverse walls, whilst the Colony Building (1894) used portal frames to provide stability.

The freedom from dependence on the external masonry to provide lateral stability created new opportunities for the treatment of the facade and architects used a variety of approaches. Ground floors were often given a light form to accommodate stores, whilst the office floors above had a traditional, heavy form. The Guaranty Building (1895) and the Stock Exchange Building (1894) both by Adler and Sullivan, and the Gage group of buildings (1896/8) by Holabird and Roche (Slide 30) are typical of this approach. One of the most simple yet successful expressions of the structural frame at the time is to be seen in the Carson Pirie Scott store by Sullivan (1904) (Slide 31). More adventurous forms, however, were possible and the bay window, supported by frames cantilevered from the spandrel girders, became a common feature, providing a means of getting light into the upper floors. This feature is perhaps best seen in the Reliance Building of 1894 (Slide 32) which used terracotta cladding over the frame to give a lightness to the form. Designed by Burnham and Root it is a notable example of the slender, glazed

skeletal building. The steel frame above the first floor was erected in little more than two weeks and the external envelope was completed within six months.

Slide 30

Slide 31

Slide 32

The greater strength of steel compared with iron enabled greater heights and longer spans to be achieved but it was relatively expensive so that it only gradually replaced wrought and cast iron, as seen in the Manhattan Building. The first all steel building was the 2nd Rand McNally Building, Chicago, built in 1889-90 and demolished in 1911.

6. DEVELOPMENTS IN DESIGN FOR STEEL FRAMED BUILDING S

In Europe the developments at the turn of the Century were less concerned with tall multi-storey buildings, but imaginative use was made of the potential for expressing the new structural material, particularly in France. Chedanne's office block at 124 Rue Reaumur, Paris (1904) (Slide 33) is perhaps the very first example of a true multi-storey facade in structural steelwork. In Belgium, too, Horta made extensive use of iron and steel, for instance, in the light wells he introduced in the deep sides of his buildings in Brussels, such as the Hotel Solay (1894). He also used it in both the elevations and the interior of the Maison du Peuple. Others used it in a highly decorative way, for instance the bridge, entrances, pavilions and canopies for the new railways in Paris and Vienna. A notable example is the Karlsplatz Station (1898) by Otto Wagner (Slide 34). The same designer combined glass and iron with considerable success in the Post Office Savings Bank, also in Vienna (1906) (Slide 35).

Slide 33

Slide 34

Slide 35

The first steel framed building of distinction in Britain was the Ritz Hotel, (Slide 36) London by Mewes and Davies and Sven Bylander. The main columns were of steel box section formed by connecting two channels lip-to-lip with cover plates. Foundations took the form of steel grillages encased in concrete, an unusual system outside the USA. The fire-proof floors were of a patented form comprising twin concrete slabs forming a floor over and a flat soffit below the steel beams. The large clear span over the restaurant necessitated the use of steel trusses. Fire protection to the steel was provided throughout by encasing in concrete or other incombustible material. The attraction of using steel was in speed of construction compared with traditional forms, even though building regulations in force at the time required the external walls to be 775mm thick. Thus, like many of its iron framed predecessors, the building displays nothing of its frame structure but instead has the appearance of loadbearing masonry.

Slide 36

Subsequent relaxations in building regulations allowed thinner wall construction and designers began to express the frame structure behind, such as at Kodak House (1911) (Slide 37) by Sir John Burnet and Heal's (1916) by Smith and Brewer, with Sven Bylander as engineer.

Slide 37

Building in the USA became even higher and architects used various design/stylistic approaches to break down their austerity such as the romantic medievalism typified by the 52 storey Woolworth Building (1913) (Slide 38) and both Gothic and Art Nouveau styles seen in the Chicago Tribunal Tower (1922) (Slide 39).

Slide 38

Slide 39

The following years saw the race to establish ever increasing height records with first the 320m high Chrysler building (Slide 40) with its famous stainless steel clad finial and the 380m high Empire State Building (1930) (Slide 41), which still holds the record for speed of construction, which at one stage reached one floor per day. The 70 storey RGA Radio Tower (Slide 42) which formed part of the Rockefeller Centre (1939) is notable since it represented the first development in which a skyscraper was planned as an integral part of a group of buildings rather than as a single structure.

Slide 40

Slide 41

Slide 42

Meanwhile, in Europe construction heights remained modest. In 1928 the Empire Theatre, Leicester Square, London providing almost 4000 seats, was constructed. Steel framing was used to span up to 36m clear over the auditorium to support a balcony with tea rooms underneath. The floor of the balcony was supported on an arrangement of raking steel beams. Other notable buildings constructed during the 1930s include de la Warr's pavilion at Bexhill-On-Sea (Slide 43), the first all-welded steel frame in Britain, and Simpson's Department Store in Piccadilly, London (Slide 44). It was the first building to have a completely clear shop front achieved by using a Vierendeel girder across the front elevation.

Slide 43

Slide 44

In France, Jean Prouvé pioneered many new applications and technical developments in the use of steelwork. Trained as a blacksmith, and specialising in metal furniture at his factory in Nantes, he collaborated with many leading architects on designs for cladding, many in cold formed steel. The Maison du Peuple, Clichy, Paris (1939) is one of his most famous works, utilizing pressed steel components throughout, not only for cladding, but also for windows, floors, partitions and staircases (Slide 45).

Slide 45

Following a lull in steel construction due to material shortages incurred as a result of the Second World War, architectural styles developed. Foremost amongst these was the influence of Mies van der Rohe and his use of a facade composed of prefabricated units and suspended in front of the structural frame. Early examples include the Illinois Institute of Technology (1950) (Slide 46), Lake Shore Drive apartments (1951) (Slide 47) and the Lever Building, NewYork (1953) (Slide 48). This new approach saved space and weight and speeded up construction, as well as allowing full visual expression to be given to glass and metal. One of the best known examples is the bronze coloured Seagram Building (1957) (Slide 49).

Slide 46

Slide 47

Slide 48

Slide 49

The evolution of form and the endeavour for increased height has continued, and these developments are chartered in Group 14.


• The use of iron and steel in the main components of building structures is relatively recent. The adoption of these new materials was spasmodic rather than a smooth sequential development.

• A historical review of the introduction of these materials may best be illustrated by the different building types - mills and industrial buildings, long span roofs and multi-storey frames.

• The first uses of iron were as secondary components - connectors, shoes and straps. Iron tie bars were incorporated in Renaissance domes. Cast iron and wrought iron were gradually adopted in structures in the 18th Century.

• Principal structural elements of iron were first introduced to achieve fire resistant construction, especially in mills.

• The developments in iron bridge construction in the 18th Century were reflected in long span roof forms.

• Over the same period iron was increasingly used in multi-storey building construction. Tall iron frame structures began to appear towards the end of the 19th Century.

• Some additional fire protection was necessary since unprotected iron beams would lose their strength at high temperatures and cast iron columns could fail when suddenly cooled by water from fire hoses.

• The introduction of the fully framed form of construction carrying the floor loads and the external wall removed the limitation of height resulting from the requirement for the wall to carry its own weight. Bracing freed the structure from dependence on external masonry to provide lateral stability. Such structures built towards the end of the 19th Century and the beginning of the 20th Century were progressively of increasing height.


1. Collins, A. R. ed., (1986) Structural Engineering - Two Centuries of British Achievement, Tarot Print, Christlehurst, Kent (1983).

2. Gloag, J. and Bridgewater, D., A History of Cast Iron in Architecture, London, 1948. 3. Lemoine, Bertrand, L'Architecture du Fer: XIXe Siecle, Paris, 1986. 4. Mainstone, R. J, Developments in Structural Form, Allen Lane 1977, London. 5. Sheppard, R., Cast Iron in Building, London 1945. 6. Jones, E, Industrial Architecture in Britain 1750-1939, London, 1985. 7. Biney, M., Great Railway Stations of Europe, Thames and Hudson, 1984. 8. Giedion, S., Space, Time and Architecture, Harvard, 1940 and 1966. 9. Russel, B., Building Systems, Industrialisation and Architecture, Wiley, 1981. 10. Guedes, P. (ed.) Macmillan Encyclopaedia of Technology. 11. Walker, D. (ed.) Great Engineers, Academy Editions, London 1987. 12. Hildelerand, G., Designing for Industry, MIT Press, 1974. 13. Ogg, A., Architecture in Steel: The Australian Context, Royal Australian Institute of Architects,

1987. 14. Strike, J., Construction into Design, Butterworth, 1991.

Lecture 1B.4.4: Historical Development of Iron and Steel in Bridges OBJECTIVE/SCOPE

To review the development of steel bridge construction, demonstrating how improvements in methods and understanding of structural behaviour have enabled greater efficiency and longer spans.

PREREQUISITES

None.

RELATED LECTURES



Lecture 1B.4.3: Historical Development of Iron and Steel in Buildings

SUMMARY

The historical development of bridges throughout the world is used to illustrate developments in structural engineering. Three categories of bridges are considered - arches, beam structures and suspension bridges. The precedence of masonry and timber construction are considered briefly, showing how these older forms have become adapted to take advantage of the characteristics of firstly iron and then steel. Significant technical innovations concerning materials, analytical methods and design concepts are highlighted. Some notable failures, and the lessons to be learned from them, are discussed.

1. INTRODUCTION

The historical development of bridges is the field which best illustrates the progress of structural engineering from ancient times up to the present century. In particular the development in steel bridges equates with the progress in structural analysis, theory of strength of materials and materials testing, since all of them were increasingly stimulated by the need for bridging larger spans and building more economically with the new construction method. Fortuitously, mechanics and mathematics had reached the threshold of modern engineering science just when the technology of constructional steelwork was being developed.

However, at the time when the new material, iron, and later steel, was ready for use in larger structures there already existed a quite highly developed technology in bridge building, namely for bridges in timber and bridges in stone. During the years 1750 - 1770 approximately, a new method of coke smelting produced larger amounts of iron at a cost which provided the basis for application of iron in engineering practice.

It is important to mention that the technologies of bridge building at that time were based on individual intuition of outstanding "masters" and on the experience passed down through the generations rather than on rules of mechanics and mathematics. The significance of preserving the knowledge of bridge building and of extending it was closely connected with military purposes and the interests of trade in ancient times. The Romans even established a separate caste - the "pontifices" (bridge makers) - who later were

raised into the rank of priests, headed by the "pontifex maximum", which was also one of the titles of the Roman emperors. Similar reasons motivated the French kings, e.g. Louis XIV, and later Napoleon, to support the new engineering schools (Ecole de Ponts et Chaussés and Ecole Polytechnique).

Thus, the building of steel bridges was founded at the beginning on the then well-tried principles and construction methods of timber and stone bridges. Stone bridges provided the arch type while wooden bridges demonstrated mainly fine-structured trusses. According to the typical material properties of cast iron -the first type of iron available - iron bridges were first built as arches. Later, when steel was available, which is capable of acting in tension, various structural systems were developed on the basis of the principles of wooden trusses. Due to the superior material properties of steel and the advantages of the new construction method, a rapid development of bridge structures led to a large variety of efficient, inventive systems for any kind of span.

In this Lecture, the history of steel bridges is subdivided according to three types of bridge:

• Arch bridges • Beam structures, including trusses, plate/box-girder bridges, and all kinds of supported bending

structures, such as cable-stayed bridges and tied arches. • Suspension bridges.

There is, of course, much overlap in chronological order concerning the three types of bridge through the period of time considered. However, this classification seems to be most appropriate to an engineer's understanding, being based on the main bearing behaviour of bridges rather than on aspects of shape or statical system.

2. ARCH BRIDGES

Arches transfer distributed vertical loads to the foundation mainly by compression. Due to the specific material properties of masonry they are basically the appropriate form of structure for stone bridges.

Such arch bridges are known to have existed in the Hellenistic period of Asia Minor. However, they reached their "flowering period" in Roman times, when the typical arch-type aqueducts were extensively used all over the Roman empire, e.g. the "Pont du Gard" near Nimes in Southern France, built in 18 B.C. (Slide 50). Up to that time arch bridges were formed in the semi-circular shape only, which did not allow spans greater than about 35 to 40 m.

Slide 50

In the Middle Ages the construction of flat arches was developed in order to build lighter bridges and larger spans. Later on, particularly in the newly founded engineering academies of France, this construction method was cultivated by using experience as well as mathematical aids. J.R. Perronet was the master of masonry bridges of that type, e.g. the "Pont de la Concorde" in Paris of 1791 (Slide 51). The technical basis for the application of iron in bridge building was therefore in place.

Slide 51

In 1779 Abraham Darby III, an English iron founder, succeeded in building the first iron bridge in Coalbrookdale. Some earlier attempts in France and England had failed because the cast iron of the time which had low tensile and flexural tensile strength, and was also brittle, had been used with inappropriate structural systems. The Coalbrookdale Bridge was constructed as an arch bridge like the examples in stone before, however, the arch was structured in 5 light ribs following the constructional principles of wooden structures. The bridge has a span of about 30 m and is still in use. Such cast iron bridges soon became common structures in Britain and were exported to other countries (Slide 52).

Slide 52

In the following years, iron casting was developed to supply different bridge building methods. Pre-fabricated, block-shaped elements were used like large "bricks" in patented iron arch bridges. The largest of these was the "Sunderland Bridge", built in England in 1796 with a span of 72 m.

Another method was developed by the German engineer Reichenbach, who used cast iron tubes for the compression member of the arch. This economical system was widely used, an excellent example being the "Pont du Caroussel" in Paris, which was built by Polonceau in 1839 with three spans of 48m each (Slide 53).

Slide 53

The largest cast iron arch ever built was the "Southwark Bridge" by John Rennie over the Thames in London (1819) with a span of 73 m (Slide 54).

Slide 54

A similar bridge, notable for its marvellous latticed design and the great name of Thomas Telford connected with it, had been built some years previously (1812) in Scotland with a span of 50 m. Other arch bridges of the same typical design were built later and can be found in many places, e.g. over the River Rhine in Germany or over the River Loire in France.

Thomas Telford (1757 - 1834) - originally being a mason - became one of the most notable engineers of his time. After educating himself in architecture he built 3 bridges over the River Severn, after which he worked for the canal companies, building about 900 miles of road and two gigantic aqueducts to carry the canals over valleys. Between 1819 and 1826 Telford built the two famous chain suspension bridges over the Menai Straits and the River Conway. Telford was made the first President of the Institution of Civil Engineers when it was founded in 1828.

The "Mississippi Bridge" in St. Louis was built in 1874 by J.B. Eads (Slide 55). He used tubular members partly of iron and partly of steel to form the latticed arch of 159 m in span. It was the first bridge he built and surprisingly became the largest arch span in the world.

Slide 55

Steel arch bridges cannot be discussed without appreciating the contribution of Gustave Eiffel, one of the greatest engineers of his century. Eiffel (1832 -1923) founded and led the "Société Eiffel", an engineering and steel fabricating company, well known throughout the world, with agencies in the Middle East, Eastern Asia and South America. Its main field of production was various kinds of steel bridges, of which the arch bridges were the most important. Eiffel also used trussed construction. He was the first engineer to develop the preparation of steelwork design up to full detailing and drawing of every element or single rivet. His first big success was the railway bridge over the Duoro in Portugal (1878) with an arch span of 160 m. His most beautiful bridge was the "Viaduc de Garabit" in the South of France, built in 1884 with a span of 165 m (Slide 56). The buildings which made him most famous are the 300 m high "Eiffel Tower" (1889) and the "Statue of Liberty" (1886).

Slide 56

With the development of steel the size of structures increased. The largest arch spans were built in the years up to 1930:

• the "Bayonne Bridge" in New Jersey by O.H. Ammann in 1931 with a span of 504 m (Slide 57)

Slide 57

• the "Sydney Harbour Bridge" by R. Freeman in 1932 with a span of 503 m (Slide 58).

Slide 58

Both bridges are two-hinged trussed arches with the deck suspended.

3. BEAM STRUCTURES INCLUDING TRUSSES AND PLATE/BOX GIRDER BRIDGES

It was indicated at the beginning of the section "Arch Bridges" that iron in the first period of bridge building could only be used in compression. It was not until more than fifty years later when larger bridge structures were built that bending structures were adopted using the newly developed wrought iron, and later steel, which were capable of acting in tension as well. At that time there already existed a highly developed technology of building such bridges in timber, in particular trusses of various shapes and systems. Since constructional steelwork at the start used a great deal of this knowledge a short overview is given below of the development of wooden bridges.

Wooden bridge structures

In Roman times (during the reigns of Caesar and Trajan) individual wooden bridges of impressive dimensions were built over the River Rhine and the Danube. Wooden bridges then became very common in the Middle Ages, although few of them have survived. The first methodical studies of statical systems were performed by the Italian architect Andrea Palladio (+ 1580), demonstrating different types of trusses and strutted frames, which were then called "Palladian bridges".

The heyday of bridge building in timber took place in the second half of the 18th century, when individual master builders like Grubenmann and Ritter in Switzerland, Gauthey in France and Wiebeking in Germany developed outstanding structures with spans up to 100 m. From that time on the development of wooden bridges moved to the USA, where - due to the lack of trained carpenters - simplified structures came into use. Standardized and prefabricated elements and simple connections were made with unskilled labour, but nevertheless produced large bridge structures, especially for the railways. The main types of bridges resulting were trestle bridges (Slide 59) and truss bridges.

Slide 59

The latter - among others - comprised patented systems like the widely used crosswise-pretensioned truss girder by Town (Slide 60). Many of the structural ideas were transferred to trussed steel bridges at the beginning. Due to the superior material behaviour of steel, wooden bridges were replaced step by step up to the end of the 19th century.

Slide 60

During the first half of the 19th century, steel bridges were frequently designed as trusses, particularly in the USA. This was mainly due to their economical load-carrying behaviour. However, in Europe this same development was interrupted by a short period, when tubular bridges were made of large plated girders.

Development of plated girders - Robert Stephenson

When in 1844 the Chester & Holyhead Railway Company decided to build a railway line from London to the Isle of Anglesey in Northern Wales, two big obstacles had to be bridged, namely the Menai Street and the River Conway. Robert Stephenson (1803 -1859), the son of the great George Stephenson, was in charge of the project. He, in contrast to his father who had been self-taught, was well educated. He

became leader of his father's locomotive factory at the age of twenty-seven and was at that time well renowned as a railway and bridge engineer in Britain.

After several studies of bridging the Menai Strait with an arch bridge or using a chain suspension bridge, which Thomas Telford had built about 20 years previously in the same place for the railway, Stephenson decided to build a bridge in the shape of two rectangular tubes (each 4,4 m wide and 9 m deep) through which the two railway tracks ran (Slide 61 and 62). He performed the design on the basis of extensive experimentation on models in the scale 1 : 6 with circular, elliptic or rectangular cross-section. The research was done in a team together with W. Fairbairn, responsible for the testing, and E.Hodgkinson, performing the theoretical work. It showed that the closely stiffened plate-girders made of wrought iron combined with the cellular upper and lower deck construction were strong enough to carry the load over the spans of 142 m without additional support by stays from the top of the piers. Such stays had originally been provided when erecting the towers, which then gave the bridge its unique appearance.

Slide 61

Slide 62

The bridge, which consisted of 4 spans of 70 + 142 + 142 + 70 m, used 10.600 tons of iron and incorporated 3,5 million rivets. It was fabricated near the site in equal pieces for each single span, and each of them was floated to the site and lifted to its final position. Both fabrication and erection were masterpieces. When the "Britannia Bridge" as it was known, was opened in 1850, Stephenson could not have known how much he had contributed to the development of plate girder construction. It was about 90 years before plate girder bridges of similar spans could be built again. The Britannia bridge carried the railway traffic well for 120 years until 1970 when it was damaged by a fire.

A second bridge of this type, but with somewhat smaller spans, was built by Stephenson over the River Conway at the same time.

Truss bridges (parallel girders)

As already mentioned the building of steel truss bridges was highly influenced by the examples of wooden trusses, built using various systems in the USA. In the first period especially, when only flat members were available, the latticed girders by Town were copied in steel, resulting in fine-mesh lattice girders since flat sections can resist compression forces only with reduced buckling length. Nevertheless, the lattice girders showed good statical behaviour and soon were built with considerable spans.

The largest beam bridge of this type in Europe, the "Dirschau Bridge" over the River Weichsel (Vistula) in Germany, was completed in 1857 (Slide 63). The single-track railway bridge was built by the great bridge engineer Karl Lentze (1801 - 1883) with six spans of 131 m each, using closely spaced lattice girders. His design was largely influenced by the Britannia bridge, showing a similar tubular cross-section as well as similar tower-like pillars. This bridge moreover shows a "speciality" of some Germany bridges, i.e. a castle-like entrance building, which was sometimes ironically criticized in other countries. Nevertheless, considerable economies in the use of steel resulted, the Dirschau bridge needing 8,3 tons of iron per metre compared with the 12,5 tons of the Britannia bridge.

Slide 63

The first iron truss bridge to be made of struts was the "Grandfey Viaduct" near Fribourg in Switzerland, opened in 1862 (Slide 64). Although similar in type to the wooden trestle viaducts in the USA (see Slide 59), it was the first true modern trussed girder with appropriate compressive members. The bridge had seven spans of 49 m and was erected by launching the girder over the high steel piers.

Slide 64

Further progress in building truss girders was encouraged by new methods of structural analysis. Karl Culmann (1821 - 1881), then a young German engineer, was sent to the USA in 1849 by the Royal Bavarian Government in order to report on the novel wooden and iron bridge types which he found there. His studies led to the development of graphic methods of structural analysis, which he published by 1860, when he was professor at the ETH Zurich. From that time a full theory existed for the design of trusses.

A typical truss bridge of that time was the Danube Bridge near Stadlau in Vienna. It was built in 1870 as a continuous beam with five spans of 80 m each. The picture (Slide 65) shows the process of launching.

Slide 65

Building truss girders was developed to perfection by G. Eiffel - as already described in the section "Arch bridges". Eiffel built a great number of truss bridges for the railway in France and Portugal; an example (Slide 66) is taken from the Beira-Alta line (1879 - 1881) in Portugal. Eiffel's largest bridge of this type was the bridge over the Tardes near Evaux, with a main span of 105 m (72 + 105 + 72 m), built in the same period.

Slide 66

Truss bridges of the parallel-girder type were built in great variety, especially for the railways in Europe, with a tendency towards simpler statical systems, e.g. the triangular truss. The Rhine Bridge near Maxau in Germany, built in 1938, is a good example (Slide 67). It is a combined railway/road bridge with spans of 175m and 117m.

Slide 67

Pauli girder, Saltash Bridge, Lohse girder

(fish-belly or parabolic girders)

The objective of obtaining an optimum distribution of the chord forces in trusses led to new shapes of girders, the parabolic-truss girder with a curved upper chord and the fish-belly type girder with both chords curved in opposite directions. The latter, called the "Pauli girder" in Germany, turned out to be very economical with chord forces being approximately constant along the length of the bridge. This system was developed by Friedrich August von Pauli (1802 - 1883), a railway engineer of the Royal Bavarian Government and later Professor at the Technical University of Munich.

The first Pauli girder, built in 1857, was the railway bridge over the Isar near Groβ hesselohe (Germany) with spans of 53 m (Slide 68). It was built under the direction of the young Heinrich Gerber (1832 -

1912), who afterwards became one of the great bridge engineers in Germany. Gerber contributed much to the design and analysis of the Pauli girder. However, his wide reputation resulted from the development of the cantilever bridge.

Slide 68

In order to summarize the great German bridge engineers of the 19th century also Johann Wilhelm Schwedler (1832 - 1912) has also to be mentioned. He contributed much to the progress of German constructional steelwork. One of his ideas was a specific parabolic truss girder, frequently used in Germany and called the "Schwedler Girder", which was designed so that none of the diagonals would be subjected to compression.

A gigantic bridge of the fish-belly type was the "Saltash Railway Bridge" near Plymouth, also known as the "Royal Albert Bridge" (Slides 69 and 70). Completed in 1859 and having two spans of 139 m each, the Saltash Bridge had a tubular upper chord with a high elliptic cross-section (5,2 m × 3,7 m), made of riveted curved plates, and a lower chord consisting of chains. Constructional difficulties prevented this type of bridge being built again. The builder was Isambard Kingdom Brunel (1806 - 1859), a renowned railway engineer in Britain and one of the most ingenious engineers of his time, whose father built the first tunnel below the River Thames in London. Brunel, after finishing his studies in France, became assistant engineer on the project of the Thames Tunnel. Although he also later built two chain suspension bridges, his greatest railway work was the Royal Albert Bridge. Brunel also designed both the first and the largest steam ships for transatlantic voyages and was also involved in the construction of many docks, piers and hospitals.

Slide 69

Slide 70

Fish-belly type girders of Pauli's design had a lot of constructional advantages and were used in German bridges again and again. For example the second "Dirschau Bridge" over the River Vistula built by J.W. Schwedler in 1891 had six spans of 131 m. The amount of structural steel used for the new bridge which carried two railway tracks, was the same as for the first bridge built in 1857 having only a single track (Slide 71).

Slide 71

A similar type of bridge was the double bow girder bridge, called "Lohse Girder" after its originator, the German bridge engineer Hermann Lohse. The structural system, somewhere in between the fish-belly type and the tied-arch type, consisted of two trussed chords connected with vertical members. The most important examples are the five Elbe Bridges near Hamburg built in the period from 1872 - 1892; one railway bridge over the Southern Elbe and a road bridge and three railway bridges over the Northern Elbe. All are of similar shape, having three or four spans of about 100 m each (Slides 72 and 73) and, again, the large entrance buildings typical for that time.

Slide 72

Slide 73

Parabolic truss systems were also widely used, particularly for railway bridges across the large rivers in Germany. The "Lek Bridge" near Culenborg in the Netherlands had the longest span of such girders for a long time. It was built in 1868, using steel for the first time in bridges, by the German engineer and fabricator J. Caspar Harkort. The truss had a span of 155 m and a depth at midspan of 20,5 m (Slide 74).

Slide 74

Cantilever bridges, Gerber beams

Nearly all bridges of the first half of the 19th century were single span beams, which means that multi-span bridges were divided into single spans on the piers. Of course, engineers of that time were aware of the beneficial statical behaviour of the continuous beam. However, they knew also of the disadvantages in relation to foundation settlements. It was the idea of the German H.Gerber to introduce hinges into continuous beams at statically favourable locations, which eliminated the drawbacks of settlements. This idea was patented in 1868 and such beams were called "Gerber beams".

Heinrich Gerber (1832 - 1912) was one of the most important bridge engineers in Germany. After his time in the Royal Bavarian Railway Authority he became the head of a significant German steelwork company and contributed much to the development of steel bridges. He was the first to introduce Wöhlers design principles for fatigue in railway bridge construction.

A special type of truss structure following Gerber's principle of hinged beams is the cantilever bridge. By making the truss girder deeper at the piers, cantilevers may be built far into the middle of the span without the need for any centring (falsework). This technique is of great importance when bridging deep or rough water.

One of the greatest cantilever bridges is the "Firth of Forth Bridge" in Scotland. When built in 1883 - 1890 with main spans of 521 m, it gained the world record for the longest span bridge (Slide 75). Some historical background of the specific design realised by the two engineers Sir John Fowler (1817 - 1898) and his partner Benjamin Baker (1840 - 1907) is given below.

Slide 75

When construction of the bridge was about to start, the design was that made by Sir Thomas Bouch, a renowned bridge engineer, who had just finished the railway bridge over the Firth of Tay with a total length of 3200 m. This was a multiple-span truss bridge with main spans of 75 m, which collapsed in a heavy storm on 27 December 1879 just when a train was crossing, causing the death of 72 people (the German poet Theodore Fontane wrote a famous poem about this accident). As a result, Thomas Bouch lost all credibility with the railway company, his successors, J Fowler and B Baker, having to illustrate the statical principles of their design to the public (Slide 76).

Slide 76

The bridge, which today is considered to be a unique and gigantic construction, is a masterpiece of engineering work. The depth of the truss above the piers is 106 m, the main tubular members are 3,7 m in diameter, and the whole bridge used 42.000 tons of steel and at times required up to 4.600 workers at the site to undertake the complex method of construction (Slide 77).

Slide 77

J Fowler was a notable civil engineer, mainly involved in railway construction. He was a pioneer of the London Underground and later elected President of the Institution of Civil Engineers. Just how much the builders of the Forth Bridge accomplished can be recognized by comparisons with the "St Lawrence Bridge" near Quebec. This cantilever bridge, very similar in type, became the longest hinged beam bridge when built in 1917, with a span of 549 m. However, although only 27 m longer in span than the Forth Bridge, it took 12 years to build, two major failures having occurred during construction, indicating that theoretical and practical limits had been reached.

The "Hooghly River Bridge" in Calcutta, built in 1940 with a span of 455 m, is the fourth largest cantilever bridge (Slide 78). Although a late example of this successful type of bridge, the design seems not to be so clear as that of its predecessors.

Slide 78

Truss Bridges in the USA

Based on a good tradition of wooden truss bridges, it was Squire Whipple who first developed the method of analysing and designing trusses made of cast and wrought iron. He was called the "Father of iron truss bridges" and built his first bridge in 1841, a bow-string type truss (parabolic girder), which was patented and successfully built many times in the years following. In 1847 he published a book on bridge building and developed the trapezoidal truss bridge, called the "Whipple-truss". Whipple built two of these bridges with spans of about 45 m for railroad use in 1852-54. These bridges have chords with forged wrought iron links, which were in later years modified step-by-step by Linville into eye-bars made of steel and accordingly allowed increased spans. The longest bridge of this type, with a main span of 155 m, was built for the railway in 1876 over the Ohio River near Cincinnati. The longest simple span truss of this time was a bow-string truss with spans of 165 m also over the Ohio River in Cincinnati, built by Bouscaren in 1888.

There were also cantilever bridges built in the USA during the period 1877 - 1889, which have main spans of 165m. They were erected by use of falsework, e.g. the "High Bridge" across the Kentucky River and the "Hudson River Bridge" at Poughkeepsie.

Plate/Box Girder Bridges

After the exceptional example of the Britannia Bridge, plate girder bridges remained within spans of about 30 m. Fresh impetus was given by the development of welding in constructional steelwork. The use of welding began in about 1925 and considerably influenced the building of steel bridges, particularly road bridges. After setbacks in the 1930's due to brittle fracture failures, a very rapid increase in the size of spans took place. A typical example of large spans is the "Rhine Bridge" in Bonn (1948) with spans of 99 + 196 + 99 m (Slide 79).

Slide 79

Tied Arches

A tied arch bridge acts like a beam structure, which is assisted in carrying load by an arch behaving similarly to a curved upper chord of a truss, while the deck girder acts like the lower chord. Arch and deck girder are simply connected by hangers and form a structure which has considerable constructional advantages compared to true trusses when bridging wide single spans or carrying heavy loads. Tied arch

bridges have been incorporated in this section because their main statical behaviour resembles beams rather than arches, e.g. transmitting vertical reactions to the abutment when subjected to vertical loads.

Such bridges were frequently used in the past, especially for heavy railway bridges. The first long span bridges were built in Hamburg over the Southern Elbe (1899) with four spans of 100 m. In 1906 - 1910 in Cologne the "Hohenzollern Bridge" was built with spans of 102 + 165 + 102 m (Slide 80). When the old Lohse girders in Hamburg had to be replaced (1915) tied arch bridges were also used (Slide 81).

Slide 80

Slide 81

Cable Stayed Bridges

Similarly, to tied arch bridges, cable stayed bridges are classified under the topic of beam structures. They actually behave like elastically supported continuous beams rather than like suspension bridges, although are often considered as being related to them. The cable stays provide a more or less elastic support at individual points along the deck girder. This arrangement allows bridges of considerable span to be built with relatively slender girders. Only vertical reactions are transmitted to the abutments as a result of vertical loading.

The cable stayed bridge was the most recently developed of all the types of bridges. It originated in Germany (about 1950) and the first bridge completed in 1957 was the "Theodor Heuss Bridge" in Düsseldorf (spans of 108 + 260 + 108 m). A great number of such bridges, mainly different in the type of pylon and the cable design, were built along the River Rhine, e.g. the harp-shaped design in Düsseldorf/Oberkassel (Slide 82) or the closely spaced, fan-shaped design in the North of Bonn Bridge (Slide 83).

Slide 82

Slide 83

4. SUSPENSION BRIDGES

The predecessors of iron and steel suspension bridges were pedestrian bridges made of rope utilising different materials during the early centuries in China, India and South America. Iron chain suspension bridges are of Chinese origin, the oldest known bridges having been built about 500 years ago. None of them were stiffened. They swayed violently under traffic and their thin decks were directly secured onto the chains.

The first proposal for a chain suspension bridge with a horizontal traffic deck suspended from three chains was published by Faustus Verantius (1551 - 1617), a Renaissance scholar, but it was not until the late 18th Century that such bridges were built (Slide 84).

Slide 84

The first of them was built by James Finley (1762 - 1828) in 1796 in the United States, followed by a large number of the same type, Finley having been granted a patent. Finley's bridges were relatively stable and could therefore be used by wheeled traffic.

Chain Suspension Bridges

The first chain bridges in Europe were erected in Great Britain.

In 1819 Samuel Brown (1776 - 1852) built the "Union Bridge" near Berwick with a span of 120 m after having invented a new type of chain, the so-called "Eye-bars". (Following this invention the fabrication of chains moved from the manufacturing of ordinary anchor cable type chains in blacksmiths' shops to wrought iron fabricators).

Brown built further chain bridges, e.g. in 1820/21 the "Trinity Pier Bridge" in Newhaven near Edinburgh (3 chainbridges in a row, each 64 m in span) and in 1822/23 the larger "Chain Pier" in Brighton, which was designed as four chain bridges of 78 m span in line. This bridge suffered from wind-induced vibrations and parts of it were destroyed twice in major storms.

It is interesting to know that, even in 1823, Marc Isambard Brunel (1769 - 1849), the builder of the Thames tunnel in London and father of the great railway engineer I K Brunel, built two chain bridges on the Isle of Réunion which were effectively stiffened against wind by additional counter-curved chains located below the bridge deck.

A milestone in bridge building was the chain bridges built by Thomas Telford, who has already been mentioned in the section on arch bridges.

The Chain Bridge over the Menai Straits in North Wales (Slide 85), being a road bridge with a free span of 177 m, was the bridge with the longest span of the time. Built in 1819 to 1826 (Telford was 60 years old when it was finished), it was an outstanding structure which also influenced Navier when working out

his theory on suspension bridges. Telford used eye-bar chains with special improved links. There were 16 chains for each of the two cables. Originally built without stiffening elements it was reinforced during the first year of use after a heavy storm had caused large deflections of about 1 m.

Slide 85

A similar bridge, but of smaller span, was built by Telford over the River Conway near Conway Castle. It should be mentioned that in the case of both the Menai Straits and the River Conway, famous railway bridges were built by Robert Stephenson, close to those of Telford, about 25 years later.

The name of another great engineer, Isambard Kingdom Brunel (1806 - 1859), is also connected with suspension bridges. Brunel, well known for his Royal Albert Bridge, a tubular-type bridge at Saltash, built the "Clifton Suspension Bridge" near Bristol (Slide 86). This chain bridge with a span of 214 m was not finished before 1864. It used the same chains as the "Hungerford Bridge" (span 206 m) in London, which had been built by Brunel in 1845.

Slide 86

Another British engineer, W T Clark, built chain bridges during this period, e.g. the "Hammersmith Bridge" in London (1827, span 122 m) and the bridge across the Danube in Budapest (1845, span 203 m).

The oldest suspension bridge in Germany was the chain bridge in Malapane (Schlesien), built in 1827 with a span of 31 m. It was followed in 1829 by the "Ludwigs Bridge" across the Regnitz in Bamberg

with a free span of 64 m. This carefully designed bridge is of some interest, since it made a deep impression on the young Johann Roebling when he was studying in Berlin. In later years he became the most important suspension bridge engineer.

Further old chain bridges, which are not referred to here in detail, were built in France, e.g. in Paris across the Seine by De Verges (1829, span 68 m) and in Langon over the Garronne by P D Martin (1831, span 80 m). Chain bridges were also built by the Czech B Schnirch in Prague (1842, span 133 m) and in Vienna (1859, span 83m).

Wire Cable Suspension Bridges

Whilst the building of chain bridges continued in Great Britain and Germany, in France, Switzerland and America wire cables began to be used, based on the experience that wires have considerably higher strength than iron chains. Following trial structures built by the French Séguin brothers, the Swiss engineer G H Dufour (1787 - 1875) and Marc Séguin (1786 - 1875) built the first wire cable suspension bridge in the world. This bridge, the "Pont St. Antoine" situated in Geneva, was, when completed in 1823, also the first permanent suspension bridge on the European continent. Six cables of 90 wires each supported the two 40 m spans.

The main problem in the manufacture of parallel wire cables is to guarantee that all wires carry the same amount of tension. While Séguin, being more an entrepreneur than an engineer, tried to achieve this by using cables of different curvatures, Dufour solved the problem by prestressing all wires so that none remained slack. This meant prestressing the cables in a special device and lifting them afterwards onto the saddles. The best solution, i.e. spinning the cables in situ wire by wire was first suggested by the French engineer L J Vicat and developed as a mechanized spinning method by J Roebling.

Although Séguin founded a bridge construction company and built more than 80 suspension bridges of about 100 m in span, the most important example of this generation of wire cable bridges was completed in 1834 by the French engineer J Chaley (1795 - 1861) in Gribourg, Switzerland. It crosses the Saane Valley in a single span of 273 m. It was called "Grand Pont Suspendu" (Slide 87) and was the longest bridge in the world until the "Ohio Bridge" in Wheeling was opened in 1849. Chaley provided 4 cables, each with 1056 wires, and prestressed them like Dufour had done before him. The cables were layed out on the bottom of the valley and lifted up to the top of the towers.

Slide 87

An interesting design was realized with a 'row' of suspension bridges crossing the Dordogne near Cubzac (Slide 88). This consisted of 5 spans, each 109 m in span, having, in addition to the main cables, separate stays which are secured on the top of a tower and land on the next tower at the height of the traffic deck. This bridge was completed in 1839 and was built by de Verges and Emil Martin.

Slide 88

After this period further development of large suspension bridges moved from Europe to the United States, partly due to the expansion of the railway to the west of the country, and also thanks to the emigration of European engineers to America and the transfer of technical knowledge. Two names dominated the major progress of this time, namely Ellet and Roebling. Whilst Ellet is thought of as a rather efficient engineer and clever entrepreneur, the Roeblings, both father and son, with their excellent scientific knowledge and technical skills, gave a major impetus to the art of building suspension bridges.

Charles Ellet (born 1810), being of poor origin, was an example of a self-made engineer. After working as assistant engineer and saving money he decided to study in Europe at the Ecole Polytechnique in Paris. He completed his studies successfully and after that travelled throughout France, Great Britain and Germany visiting the newest bridges and engineering works. On returning to the United States he became very active as an entrepreneur, working on projects for large suspension bridges and proposing them efficiently. During this time he came in contact with J A Roebling, who suggested cooperation, but was rejected, this being the beginning of their subsequent rivalry for life.

After building a number of successful bridges, the biggest success of Ellet was the suspension bridge over the Ohio near Wheeling. Finished in 1849 with a free span of 308m it was the longest bridge of that time. The two cables consisted of 6 ropes each, each of them comprising 550 wires, grouped side by side so that, if strengthening the bridge became necessary for railway operation, further ropes could be added. Before it could be demolished (having insufficient clearance for steamboats), it was destroyed in 1855 in a heavy storm. Six years later it was rebuilt by Roebling.

Railway Suspension Bridges

Before discussing the Roeblings in detail, some remarks should be made concerning the use of suspension bridges for railways. The first attempt was made in 1830 by building a chain bridge over the River Tees near Stockton for an extension of the Stockton-Darlington line. The free span was 86 m, the calculated live load 150 tons, but disappointingly under less than half of the load the deflections were intolerably high. This behaviour accounted for the ill repute of such bridges for bridging railways. However, the

suspension bridge engineers in the United States, like Ellet and Roebling, were optimistic or even convinced that suspension bridges for railways could be achieved. Their first major test came with the crossing of the Niagara gorge (see below). Following this, the Brooklyn Bridge was also designed to carry railways. Very few railway suspension bridges have been built since then. An exception was the railway chain bridge built in Vienna by Schnirch across the Danube canal with a span of 83 m (1859).

The Roeblings

The main development of suspension bridges up until the work of the Roeblings had been carried out by British and French engineers. Johann August Roebling (1806 - 1869) was born in Thüringen, Germany, studied at the then famous engineering school, the "Royal Polytechnic Institute" in Berlin, and emigrated in 1831 to the United States. There he became one of the greatest bridge building engineers of that continent as well as the leading fabricator of wire rope. Working first as a surveyor for canal companies he invented machines for manufacturing ropes from wires and then developed an efficient wire rope firm, which later, under the management of his sons, had 8000 employees. Between 1844 and 1850 he built 5 Cabak crossings over Rivers, i.e. aqueducts, as well as one road bridge which were all supported by wire cables. These aqueducts, carrying the high mass of water of the canal in wooden troughs made him a notable engineer. Some of them are still in use even today after having been converted into road bridges.

He developed a mechanized cable spinning method in which wires were carried by a wheel back and forth over the towers and anchorages. Using this method the requirement that all wires should be under the same amount of tension could be realized in a natural way giving every wire the same curvature (sag). Modern methods of manufacturing suspension cables are, in principle, still the same. Some of the operations executed manually in Roebling's time have since been mechanized.

The idea of a railway crossing the Niagara gorge (Slide 89) near the falls was a great challenge to American and European bridge builders. While European engineers like Samuel Brown and Robert Stephenson thought a free span of 250 m for the load of railway traffic to be too risky or even impossible, the Americans Ellet, Roebling, Serrel and Keefer - all being competitors - applied for the project. The first to be successful in winning the contract was Ellet in 1847, but he only built a temporary pedestrian bridge and failed to realize a railway bridge. The next was Roebling in 1851 and he succeeded, building a double-deck bridge for railway and road traffic. The girder was a wooden Howe truss and the four cables consisted of 3640 wires each. When the bridge was opened in 1855, being the first railway bridge of a span of 250m, it made Roebling a very respected engineer. Serrel and Keefer also built suspension bridges across the Niagara, the first one a road bridge (1851 with 318m span, destroyed in 1861 in a storm), the second a footbridge very close to the falls (1868, span 388 m) called the "Honeymoon Bridge", which was also destroyed in a storm in 1889.

Slide 89

The main innovation of Roeblings work was his efficient conceptual design, which allowed for the effect of storms through stiffening by diagonal stays as well as additional stays beneath the roadway. Roebling was also the first to build suspension bridges with systematic rigidity of the deck girder. He published his theories stressing the importance of considering wind effects in the design. It is perhaps surprising that many engineers later forgot the importance of wind effects, culminating in the famous accident at Tacoma Narrows of 1940 (see later).

In the period 1857 - 1866 Roebling built the "Allegheny suspension bridge" in Pittsburgh and then the large "Ohio River Bridge" in Cincinnati with a span of 322 m, which made it the longest in the world when completed in 1866. In this bridge, wrought iron beams and trusses were used for the deck girders. During construction of both bridges Roebling's son, Washington A Roebling (1837-1926), worked as assistant to his father. The Roeblings dream, or even obsession, was to build a bridge over the East River, between Brooklyn and New York. Their idea was for a suspension bridge for railway and road traffic with a span of 486 m. But J A Roebling was not able to realize the project himself due to a mortal accident on site during surveying work, only 3 years after winning the contract. His son took over his position, but during the work in the pneumatic caissons for the foundation of the towers he suffered a serious collapse from caisson disease. From that time on he was an invalid, bound to his bed and suffering from a nervous disorder. He ran the project from his sickroom, located close to the site, watching the progress of the work through a field glass from his window.

His wife, Emily Warren Roebling dedicated her life to the bridge, became his assistant and kept contact with the workers and fellow engineers. When the Brooklyn or East River Bridge (Slides 90 and 91) opened in 1883 it was a masterpiece of engineering work, the largest bridge in the world. The towers, built of masonry, were 107 m in height; the anchor blocks 60.000 tons in weight each; the 4 cables 40cm in diameter, consisting of 5358 wires each; stiffened by a deep trussed deck girder and a large number of diagonal stays.

Slide 90

Slide 91

After more than 100 years since it was opened, the Brooklyn Bridge is still in use.

Increasing the Spans

After the Brooklyn bridge, which reached roughly 500m in span, the spans of suspension bridges still continued to increase in size. Fifty years later the previous record span had doubled.

In 1931 the "George Washington Bridge" (Slide 92) in New York was the first structure to span over 1000 m. Othmar H Amman, an emigrated Swiss engineer who became one of the great bridge builders in the United States, used 4 cables of 91 cm diameter and over 20.000 wires each. The bridge carried the greatest live load of any bridge, consisting of two traffic decks and 14 lanes and has a span of 1067 m.

Slide 92

Certainly the most famous of all suspension bridges is the "Golden Gate Bridge" (Slide 93) across the entrance to San Francisco. It was built by Joseph Strauss in 1937 with a span of 1281m. Besides the marvellous shape of the bridge it is interesting to note that the colour of the bridge was carefully selected, resulting in "International orange". Any attempt to change it has been fiercely opposed by the people of San Francisco.

Slide 93

The "Tacoma Narrows Bridge" (Slide 94) near Seattle, with a then average span of 853 m, sadly became renowned when it collapsed in 1940 under wind. The failure was recorded on film. Engineers, dedicated to the opportunities of statical calculations, made continual efforts in building more economical and more slender structures, not being aware of the lectures Roebling had given before on stiffening bridges against wind. The Tacoma bridge was caused to oscillate by wind, although the statical theories - as then known - had been correctly applied. Design methods were revised after this accident and, as a result, new directions developed in the design of suspension bridges:

Slide 94

• One direction was taken by O Amman in the United States when designing the "Verrazano Narrows Bridge" (Slide 95), the largest span of that time at 1298 m, crossing the entrance to New York Harbour. He chose a very stiff box girder to withstand torsional vibrations due to the dynamic influences of wind.

• Another direction was taken in Europe, where profound knowledge of aerodynamic problems led to the use of decks similar in shape to the wings of aeroplanes. The newest bridges in Great Britain have been built in this way and one of them - the "Humber Bridge" - established the world record for free spans of 1410 m (Slide 96).

Slide 95

Slide 96


• Early iron bridge construction assumed similar forms to those traditionally used for masonry and timber bridge construction.

• Significant developments in iron and subsequently steel bridge construction have enabled longer spans, improved efficiency and greater elegance.

• These developments are associated with an improved understanding of structural behaviour and better material properties.

• Also critical in this development has been the engineers' ability to create new design concepts and to perform sophisticated analyses.

• Developments in bridge construction have not been without failures.


1. Robins, F. W., The Story of the Bridge, Birmingham, Cornish 1948 2. James, J. G., The Evolution of Iron Bridge Trusses to 1850, Transactions of New Common

Society, Vol 52 (1980-81), pp 67-101. 3. Walker, J. G., Great Engineers, Academy Editions, London 1987

Lecture 1B.5.1: Introduction to the Design of Simple Industrial Buildings OBJECTIVE/SCOPE

To describe the reasons for the use of steel and to present common forms of structure for industrial buildings.

PREREQUISITES

None.

RELATED LECTURES

Lecture 1A.1: European Construction Industry



Lecture 7.12: Trusses and Lattice Girders

Lecture 14.1.1: Single Storey Buildings: Introduction and Primary Structure

Lecture 14.1.2: Single Storey Buildings: Envelope and Secondary Structure

Lecture 14.2: Analysis of Portal Frames: Introduction and Elastic Analysis

Lecture 14.3: Analysis of Portal Frames: Plastic Analysis

SUMMARY

The reasons for the wide use of steel for industrial buildings are discussed. The advantages of steel include its high strength-to-weight ratio, speed of erection and ease of extension. Steel is used not only for members but also for cladding.

Common types of structure are described. These types include portal frame, lattice girder and truss construction. It is shown that overall stability is easily achieved. The wide variety of sections used in industrial buildings is presented. Possible approaches to global analysis are identified.

1. TYPES OF INDUSTRIAL BUILDING

A wide variety of building types exists, ranging from major structures, such as power stations and process plants, to small manufacturing units for high quality goods.

The most common type is the simple rectangular structure (Figure 1), typically single-storey, which provides a weatherproof and environmentally comfortable space for carrying out manufacturing or for storage. First cost is always an overriding consideration, but within a reasonable budget a building of good appearance with moderate maintenance requirements can be achieved. While ease of extension and

flexibility are desirable, first cost usually limits the provisions which can be usefully included in the design for these potential requirements. Although savings in the cost of specific future modifications can be achieved by suitable provisions, for example by avoiding the use of special gable frames (Figure 2), changes in manufacturing processes or building use may vary the modifications required.

When, for reasons of prestige, the budget is more liberal, a complex plan shape or unusual structural arrangement may provide a building of architectural significance.

While many features are common to all industrial buildings, this lecture deals mainly with single-storey buildings of straightforward construction and shape.

2. STRUCTURAL STEEL FOR INDUSTRIAL BUILDINGS

Compared to other materials, particularly reinforced or prestressed concrete, steel has major advantages. Its high strength-to-weight ratio and its high tensile and compressive strength enable steel buildings to be of relatively light construction. Steel is therefore the most suitable material for long-span roofs, where self-weight is of prime importance. Steel buildings can also be modified for extension or change of use due to the ease with which steel sections can be connected to existing work.

Not only is steel a versatile material for the structure of a building, but a wide variety of cladding has been devised utilising the strength developed by folding thin sheets into profiled form (Figure 3). Insulated cladding systems with special coatings are now widely used for roofing and sidewall cladding. They have good appearance and durability, and are capable of being speedily fixed into position.

The structure of a steel building, especially of an industrial building, is quickly erected and clad, providing a weatherproof envelope which enables the floor and installation of services and internal finishes to proceed at an early stage. Since the construction schedule is always tied to the earliest handover date fixed by production planning, time saved in construction is usually very valuable.

In a dry closed environment steel does not rust, and protection against corrosion is needed only for the erection period. For other environments protection systems are available, which, depending on cost and suitable maintenance, prevent corrosion adequately.

Single storey industrial buildings are usually exempt from structural fire protection requirements. Spread of fire beyond the boundary of the building must not occur as a result of collapse of the structure. This requirement can be met by the provision of fire walls and through the restraint which arises in practice between the bases and the columns which they support.

3. CHOICE OF INDUSTRIAL BUILDING

A prospective owner may have a fully detailed design brief derived from the construction of industrial plants elsewhere. More usually the owner is assisted in the choice of a suitable building by the completion

of a detailed list of requirements so that a design brief can be prepared. Initial options in respect of preferred location, site acquisition and environmental needs must first be decided. Then main dimensions, process operation, plant layout, foundation needs, handling systems, daylighting, environmental control, service routes, staffing level and access all require definition.

The preliminary selection must be made between a building specially designed for the owner, a new factory largely built of standard structural components, or the adaptation of an existing building. The latter may be either an advance unit built as a speculative development, or a unit which has been vacated.

The location of internal columns and the internal headroom are always important, and consideration of these requirements alone may determine the choice. The advantage of freedom to plan the building to suit requirements closely and allow for future development is very valuable. However, unless there are exceptional reasons such as permanence of specific use, it is unwise to design an industrial building exclusively for a single process, since special features appropriate to the process may make redevelopment difficult.

4. SHAPES OF INDUSTRIAL BUILDINGS

Because of its economy, the most widely used building shape is the pin-based single or multi-bay pitched roof portal frame, typically of 20-30m span at 6m centres (Figure 4). Hot-rolled I, welded or cold-formed sections are usually used for the members.

During recent years an increasing use of welded sections has occurred. This increase is the result of progress achieved in making welding automatic and the ability to adapt the cross-section to the internal forces.

Since internal columns sterilise an appreciable space around them, their spacing may be increased by using spine I-beams to support the portal rafters. For this type of roof the cladding is usually insulated metal decking, which may also be used for the upper sidewalls. Daylight is provided by profiled translucent sheeting in the roof.

When hot-rolled sections are used, haunches (Figure 5) are usually provided at the eaves and the ridge. These haunches deepen the overall section, thereby reducing bolt forces. By extending the haunched regions along the rafter the frame is also strengthened and stiffened.

Lattice girders (Figure 6) are lighter than portal frame rafters for wider spans, but the additional workmanship increases fabrication costs. Based on structural requirements alone, lattice systems are likely to be cost-effective for spans above 20m. Roof trusses may also be used for structures which support heavy cranes (Figure 7).

A wide variety of structural sections may be used for lattice girders and roof trusses, including single angles, angles back- to-back, tees, H-sections or hollow sections (Figure 8). For light loading, cold-formed sections may be used as booms, with reinforcing bars as the web members (Figure 9).

The disadvantages of multi-bay pitched roofs are that internal gutters and rainwater disposal are required, which are a possible source of leaks, and access to plant mounted externally on the roof is difficult.

The most versatile roof shape is the nominally flat roof, covered with an insulated membrane on metal decking (Figure 10). This shape allows wide freedom in plan form, and eliminates the need for internal gutters, although some internal rainwater disposal may be necessary if the extent of the roof is large. The mounting and weather protection of external plant on the roof is simply achieved, and access can readily be provided.

Flat roofs can be supported by rolled or cold-formed purlins on main I-beams or lattice girders. For smaller structures the deck may span directly from one frame to another, without the need for purlins.

When services are extensive and there are many external plant units on the roof, castellated beams or double-layer grid space frames (Figures 11 and 12) can be very suitable for flat roofs. The two-way grid distributes local loads better than any other structural form. The support for the roof deck is provided directly by the top layer and support for the services by the bottom layer of the grid. Castellated beams have a much higher moment of resistance than I-beams.

The provision of daylighting in flat roofs is expensive, since either dome or monitor lights must be used. Flat roofs are most common for industries where daylighting requirements are minimal.

5. STABILITY OF INDUSTRIAL BUILDINGS

It is essential to ascertain the loads applied to the structure and to determine the load paths from the cladding to the purlins and side rails, through the main frames to the foundations. The loads may arise from dead load, wind load and snow load, and sometimes from cranes or impact caused by fork-lift trucks.

The overall resistance of simple single-storey industrial buildings to horizontal loading is usually easy to achieve. One of the attractions of portal frame buildings is that in-plane stability follows from the rigidity of the frame connections. Stabilising bracing between the portals is therefore only required in line with corresponding rafter bracing in the roof plane.

For short buildings, bracing in one end bay may be sufficient. For longer buildings, bracing of two or more bays may be necessary.

The rafter bracing itself provides restraint to the heads of the gable stanchions. The braced end bays provide anchor points to which the longitudinal rafter stabilising ties, which are usually the purlins, are attached. During erection, bracing facilitates plumbing and squaring of the building, as well as providing essential stability.

For frames with lattice girders (Figure 6), in-plane stability can be provided by connecting both top and bottom booms to the column.

If the building has roof trusses (Figure 7), or if only the top booms of the lattice girders are connected to the column (Figure 13), the frame is effectively pinned at eaves level. To provide in-plane stability, either

the column bases should be fixed or longitudinal girders should be provided in the plane of the roof (Figure 14). These girders span between the gable ends, which must be braced appropriately. If the building is long, or is divided by expansion joints, longitudinal bracing may not be practicable and the columns must have fixed bases.

Buildings using lattice girders or truss roofs also need bracing to provide longitudinal stability.

Bracing members for industrial buildings commonly use circular hollow sections, rods or angles.

When cranage is provided the stability requirements need further examination, since longitudinal and transverse surge from the crane increases the forces in the bracing systems.

6. GLOBAL ANALYSIS

The structure may be treated either as a 2-D or 3-D system.

Bracing systems are analysed as if pin-jointed. When cross-bracing is used, for example in vertical bracing, only the members in tension are assumed to be effective (compression members are assumed ineffective because of buckling).

The choice of the method of global analysis, either plastic or elastic, of portal frames at the ultimate limit states depends on the class of the cross-section.

An example of the plastic collapse mechanism of a frame with haunches is given in Figure 15. Buildings with cranes should always be analyzed elastically. Elastic analysis should always be used to determine deflections under service loading.


• Steel construction is widely used for industrial buildings, including structural members (like frames, purlins, side rails) and cladding systems.

• Overall stability is obtained from the rigidity of connections and the use of bracing systems. • The buildings may be analyzed using 2-D or 3-D modelling and elastic or plastic analysis,

depending on their cross-sections. • A wide variety of hot-rolled shapes are available for structural members. More flexibility can be

obtained using welded sections. Purlins and side rails may be formed from cold-rolled sections.

Lecture 1B.5.2: Introduction to the Design of Special Industrial Buildings

OBJECTIVE/SCOPE

To outline the principal features of the design of special industrial buildings.

PREREQUISITES

None.

RELATED LECTURES

Lecture 1B.5.1: Introduction to the Design of Simple Industrial Buildings

SUMMARY

Special industrial buildings are of two kinds - those which are of unusual construction and those which are designed for a special industry. Several features, such as handling methods, maintenance and fire protection, are briefly discussed. Examples of special buildings, e.g. power stations, hangers, are presented.

1. TYPES OF SPECIAL INDUSTRIAL BUILDINGS

Special industrial buildings are of two kinds - those which are of unusual construction and those which are designed for a special industry. The main characteristic of such buildings is that they are invariably designed for a particular purpose or process, and are consequently virtually impossible to adapt for another kind of use.

Among the former are industrial buildings which, for reasons of prestige rather than economy, utilise unusual structural forms which provide architectural expression and thereby contribute to the visual quality of the building. Because buildings of this kind are unique they cannot be considered generically. Some examples are briefly described later in this lecture.

Among buildings designed for specific industries are heavy engineering works, aircraft hangars, power stations, process plants, steel rolling mills and breweries. Many of these buildings have similar features which are considered in principle below.

2. HANDLING METHODS

Overhead cranes with capacities of 10 tonnes and more are a characteristic of heavy engineering works and power stations. They require the support of compound columns and runway beams to carry the vertical and surge loads (Figure 1). Light overhead cranes with capacities of 1 to 5 tonnes, are a characteristic of aircraft hangars and light industries. They can be attached to the roof structure and be designed for multiple supports for wide coverage, or they can be arranged to transfer laterally from bay to bay (Figure 2). Roof flexibility may become important for roof-mounted cranes used for assembly.

Some years ago, so-called NoRail cranes were developed.

The NoRail crane concept inverts the overhead crane principle. Short rails are mounted in the endtrucks of the crane. These rails run along a series of stationary wheels. The rails are designed to be somewhat longer than the maximum distance between three adjacent support points, so that the crane is always supported by at least two wheels on each side (Figure 3). As a result of this design, the long conventional crane track becomes superfluous. The benefits of this innovative design arise both in cost savings (up to 20%) on the steel structure of the building and in material handling. Crane travel "tracks" that cross each other are feasible.

Conveyors can be either floor or roof mounted. Conveyors for assembly purposes may carry appreciable weights, and are of necessity suspended from the roof (Figure 4). Power roller conveyors are also used for transport of bulky items and are usually floor mounted.

As a result of advances in design, motorised floor transport vehicles including fork-lift and pallet trucks are now very common. The main influence they have on design is on the floor quality and on headroom.

Automated pallet stacking by fork-lift trucks of specialised design may require very stringent control of fabrication and erection of the stacking racks. The racks may be incorporated in the structure of the building (Figure 5).

3. DAYLIGHTING

Few industries now have particular needs in respect of daylighting, since shift work is often provided for. Sidewall and roof daylighting is usually described as a percentage of the plan area, 5% giving sufficient light for bulk storage, 20% for a working process. Since artificial lighting is usually employed to establish a consistent high level of illumination, daylighting may be provided for visual comfort or for architectural effect.

4. SERVICES

The amount of services can vary in different parts of a building, from an exacting standard of air conditioning appropriate to a "clean room" to extensive process ductwork. The support and passage of services can be facilitated or hindered by the roof construction (Figure 6). The heating of high single-storey structures is always a problem, particularly when fire safety places stringent control on the temperature of the heat source. Inevitably provisions for cranage, lighting, heating and services such as air and electric power, will conflict. They each influence the structural design. Sometimes, if services are particularly extensive, it is advantageous to use a structural form which provides abundant support for services.

5. SPECIAL ROOF LOADING

Whilst it is usual in advanced factory units to allow in the design of the roof a nominal overall loading for services and a single point load on the main members, this provision may not be sufficient for special buildings. Roof loading may be determined by provision for future developments of the process for which the building is designed, or for developments in handling methods or access platforms designed for improved productivity. These provisions may cause major loads on the roof. Whilst it is not possible to take into account every possible development which can influence the building design without incurring large additional cost, it is much cheaper to incorporate surplus strength in a building at the design stage than to add additional strength after completion, particularly if intensive use of the building would conflict with the strengthening operation. The ability of the structure to laterally distribute local loads may influence the choice of structure. Space frames, for example, have exceptional capabilities in this respect.

6. MAINTENANCE

Every material used in construction has a limited life, which can usually be extended by appropriate maintenance. Maintenance is likely to be particularly important in special buildings. The design of the building should allow suitable access for the maintenance required. Maintenance may conflict with the planned usage of the building, which can easily occur if usage is intensive, as, for instance, if maintenance requires dismantling or opening up, or if radiography requires areas cleared for safety.

Roof maintenance is particularly important. The possible results of overflow due to rainwater outlets being blocked, either by process emissions or by snow or hail needs to be considered in assessing the merits of the roof design, the routes for rainwater disposal and the maintenance necessary. The deterioration of the roof covering due to weather or to aggressive effluent also needs consideration.

7. FIRE PROTECTION

Due to the characteristics of the process to be carried out in a special building, it may require exceptional measures in respect of fire and explosion prevention, and in fire protection and damage limitation. Sprinkler installations of exceptional capacity may be required, as well as carbon-dioxide injection.

Dust explosion is a risk in processes dependent on the transport of finely divided powders by conveyor or air duct. Controlling the results of an explosion is often achieved by strategically placed blow-out panels. Gas explosions can be far more destructive and difficult to control.

8. SOME EXAMPLES OF SPECIAL BUILDINGS

8.1 Coal-Fired Power Stations

A typical medium-sized power station (Figures 7 and 8) consists of a 38,6m span turbine hall, flanking a 13m span bunker bay beside a 31,5m span boiler house and 12m wide air heater building. The height of the turbine hall is typically 30m, determined by the servicing requirements of the turbines and generators. The height of the bunker bay, which stores several hours fuel, and that of the boiler house are similar, determined by the height of the boiler and the size of the fuel mill below, and is typically 60m. The length of the building depends on the number of generators installed, each having its own boiler.

This type of power station is constructed almost entirely of structural steelwork and steel cladding. Steel construction is chosen because the completion of the boiler house is always on the critical path of the execution schedule. The execution of the boiler frame, designed to suit the boiler and from which the boiler is suspended, is central to the schedule. The stanchions of the boiler frame, often six in number, are typically compound H-section, carrying up to 1000 tonnes each, and the boiler is suspended from heavy plate girders spanning across the stanchions. The external steelwork to the boiler house is relatively light, being mainly supported by the boiler frame which also braces the building.

In the bunker bay, which is also a steel structure, are large feed bunkers of 600 tonnes capacity constructed of steel plate, supported at high level, to which fuel is supplied by conveyors. There is a fire and explosion hazard in the feed conveyors and the ductwork connecting the bunker to the fuel mill and the latter to the boiler. Sprinkler and carbon-dioxide fire protection is therefore required in this part of the plant, and fire protection is also applied to the steelwork.

In the turbine hall the generator sets are supported 10m above floor with condensers fitted below. Due to the weight of the generator sets the supporting structure, which is usually of steel but may be of concrete, is of heavy construction. To carry out maintenance of the generator sets a 100 tonne overhead crane travelling the length of the hall is provided, requiring heavy compound sidewall stanchions to support the runway beams. The roof structure is of light lattice girders except where additional strength is required to facilitate the installation of the crane.

Provision for extension of the turbine hall can be made, but extension of the boiler house depends on the choice of boiler, so that the ease of joining to existing steelwork has to be relied upon.

Maintenance of the generating plant is an important consideration in the design of a power station. Maintenance of the building is reasonably straightforward, since generation does not create aggressive conditions or waste. Corrosion is not a major problem, so that it is adequate to shot-blast and coat the steelwork.

The construction of power stations of this type displays the versatility of steel, its use varying from heavy steelwork for the support of plant to light roof steelwork and sheeting. Allied to this versatility is speed of execution on site, which off-site fabrication allows. It is therefore understandable that steel is used almost exclusively in this field of application.

8.2 Aircraft Maintenance Hangar

A typical hangar bay for the maintenance of Boeing 747 aircraft (Figures 9 - 11) is 76m wide and 97,5m long, and the hangar may consist of one, two or three bays. The maximum clear height is usually 23,5m to allow clearance over the 20m high tail fin of the aircraft, but only 17m is required over the body and main wings. The roof may therefore have two levels, the height in the tail area being 23,5m, the remaining area 17m. The two-level roof restricts the attitude of the aircraft to nose-first, whereas a full-height hangar allows either nose-first or tail-first attitude. At the rear of the hangar is the 2-3 storey workshop and administration block, 10m deep and the same width as the hangar. The roof slope is usually small to avoid excessive height, utilising either an insulated roof membrane on metal decking or insulated two-layer cladding. The roof structure is usually comprised of lattice trusses, girder or portal frames, but double-layer grid space frames have also been used.

The main door is usually 21m high, and can be a sliding-folding or slab-sliding design. The full opening width required is 80m. If bunching space for the doors overlaps the door opening the bay width is increased correspondingly. Some hangar doors are only 14m high with a 7m high tail gate, or they may have a vertically folding 21m high centre section.

Whilst some smaller hangars have been constructed in prestressed concrete, virtually all are now constructed in structural steelwork with insulated steel cladding.

Hangars are specialised for maintenance of one type of aircraft or a mix of types. Access to an aircraft, because of its shape and size, is a problem which is best solved by specially designed docking tailored to suit the particular aircraft. This arrangement enables a large workforce to carry out maintenance. Typically the docking consists of main wing docks, a tail dock and body dock. They are moved into place after the aircraft has been placed in a fixed position. Since aircraft are jacked up 1,5m for landing gear overhaul, it is usually necessary for the docks to have vertical adjustment. The use of wheel pits can make jacking unnecessary, but these add considerably to cost as well as adding to specialisation.

Unless they can be moved out of the hangar, docks occupy a large amount of floor space. They obstruct the placing and maintenance of other types of aircraft when not in use. Consequently tail docks and body

docks are sometimes suspended from the hangar roof. Since tail docks weigh 12-50 tonnes and body docks 50-100 tonnes, provision for them must be incorporated in the roof design.

Hangars are usually provided with light overhead cranes covering the full area. They are used to handle dismantled parts up to 1 tonne weight. Isolated engine hoists up to 10 tonne capacity may also be provided. Alternatively the overhead cranes may be of 10 tonne capacity. Conflict can arise between cranes and suspended docking. If a two-level roof is adopted, separate cranage is required in the tail bay.

Electric power, air and other services may be from roof-mounted motorised reels or in the floor. Heating is by embedded floor coils or high-power blowers suspended from the floor. Blowers are large units appropriate to the height of the hangar. Sprinklers may be installed, depending on the extent of the maintenance carried out and the safety procedures adopted regarding on-board fuel.

Except for roof maintenance, the maintenance requirements for a hangar are usually slight, since aggressive emissions are confined to drainage from the hangar floor where painting is carried out, or from cleaning or chemical process shops. Due to the large roof area and its height, and to the characteristically exposed environment of an airport, storm damage is always possible. Roof leaks can have very serious consequences, because of the high value of aircraft parts.

Developments in aircraft design and increased competition for contract maintenance make it necessary to allow for modifications to a hangar. The introduction of the 747 type and other wide-body aircraft compelled the extension of many of the hangers in use at that time. However, the intensive usage of a hangar and the strict fire and safety regulations applied when aircraft are inside makes modification difficult to carry out. Flexibility therefore needs to be allowed for at the design stage.

The superiority of structural steelwork for aircraft hangars is now well established. The speed of construction, suitability for large-span roofs, versatility for the mounting of various services and docks, and the adaptability for future development virtually exclude other structural materials.

8.3 Milk Powder Plant

A typical milk powder plant (Figures 12 - 14) consists of a spray-drier tower 18m by 17m by 32m high with an external boiler house, a silo and packing plant annex 16m by 18m, and a storage warehouse for packaged powder 54m by 54m with 7m clear height for fork-lift transport and stacking.

The tower and annex are framed in structural steelwork, with composite concrete floor and steel cladding. The warehouse typically has multi-bay short-span portal frames carrying pressed steel purlins and asbestos cement or single-skin metal cladding.

The spray drier is a 10m dia. stainless steel drum 14m high, supported at several floors. Milk and hot air are injected at the top, and the dry milk powder collects in the hopper bottom. From there it is conveyed to the silos of the packing plant. The floors are lightly loaded except for ancillary plant and the spray drier, which in operation weighs 60 tonnes.

There is an appreciable explosion risk from the finely divided milk powder. Strong explosion ducting with an exterior blow-out panel, intended to control the direction and result of an explosion, is incorporated in the drier, and provision for this facility is made in the tower steelwork.

The large amount of air injected in the process requires outlet cyclones to extract milk powder from the exhaust air. Even with regular maintenance, cyclones are never 100% efficient, so that some powder, which can accumulate quickly, escapes. Deposits of powder can cause problems with roof drainage, which therefore requires appropriate design. Milk powder contains lactic acid which is moderately aggressive particularly to flat roof coatings such as asphalt and felt. Consideration of the durability of the roof is therefore required.

Internally a biologically clean environment is required in order that the plant complies with process regulations. Easily cleanable surfaces are required internally. This requirement is best met by high quality internal sheeting. Avoidance of crevices which can cause lodgement of material affects the choice and detailing of any steelwork exposed internally.

Competition in milk powder production requires that first cost and running cost are carefully controlled. Since development of driers occurs, a change of drier may be necessary involving major alterations to the tower. The use of structural steelwork and cladding facilitates cost control in both construction and modification.

8.4 Industrial Complex

Some major industrial projects provide both the scale and the opportunity for adopting unusual structural forms which have particular advantages. A good example of an unusual structure form is the Renault Parts Distribution Centre in Swindon (Figures 15 - 17).

The requirement was for a single-storey building of 25.000sq.m containing a warehouse, training school, showroom and office, with provision for 50% expansion. To suit the storage arrangements for the warehouse a 24m x 24m bay was adopted, with 8m internal height, with 2,8% roof lighting and sidewall glazing in some areas. The main area is 4 bays wide and 9 bays long, with an additional 6 bays at one end.

The structure consists of skeleton portal frames on both rectangular and diagonal axes. The main verticals are 16m high 457mm dia. circular hollow sections with rod stiffeners. The roof members are simple trusses formed from shaped I-beams cambered 1,4m stiffened on the underside with rod bracing and short tubular verticals. Continuity between the main verticals and the trusses is established by rod bracing connecting the heads of the main verticals to the quarter-points of the trusses. Whilst the internal verticals are balanced by trusses on each side, the perimeter verticals, which have transverse and diagonal trusses on one side only, are balanced by ground anchors bracing short beam members connected to the verticals at the same level as the trusses.

Macalloy bars are used for the rod stiffeners to the main verticals, and S355 steel is used for the main rod bracing. The rods are connected to the main verticals by purpose-made cast-iron eyes pinned to lugs welded to the 457mm dia. hollow sections, and to the trusses through sleeves set into the beam sections.

In each bay the trusses are cambered to a central 4m x 4m dome rooflight. The roofing consists of an insulated membrane on metal decking, which is carried on purlins between the trusses. Valleys formed by the cambered trusses are drained by downpipes incorporated in the main verticals. Both main vertical and bracing rods pass through the roof covering.

The overall appearance is unusual, resembling a large marquee due to the tent-like profiles of the cambered trusses and the main verticals and bracing rods protruding through the roof.


• Special structures are needed for some industries. They may also be provided for reasons of prestige.

• Cranes and conveyors carry appreciable weights and may be suspended from the roof. • If services are extensive, it is advantageous to use a structural form which provides abundant

support. • Speed of construction, suitability for large spans, versatility for the mounting of services and

adaptability all favour the use of structural steelwork for industrial buildings.

Lecture 1B.6.1: Introduction to the Design of Steel and Composite Bridges: Part 1

OBJECTIVE/SCOPE

To introduce steel and composite bridges. To discuss bridge components and structural systems. To describe the common types of steel bridge - plate girder, box girder and truss girder bridges.

PREREQUISITES

None.

RELATED LECTURES


Lectures 15B: Structural Systems: Bridges

SUMMARY

The fundamentals of bridges are described. The basic components of a bridge structure are given and the types of bridge structural systems are discussed in the context of their uses. General aspects and deck systems of steel bridges are described prior to discussion of plate girder, box girder and truss girder bridges.

1. FUNDAMENTALS

Bridges have been built by man in order to overcome obstacles to travel caused by, for example, straits, rivers, valleys or existing roads. The purpose of a bridge is to carry a service such as a roadway or a railway.

Bridges play an outstanding role in structural engineering, deserving the denomination of "ouvrages d'art" in latin languages.

The choice between a steel bridge and a concrete bridge (reinforced concrete or prestressed concrete) is a basic decision to be taken at a preliminary design stage. Several factors influence this decision, for example:

• spans required • execution processes • local conditions • foundation constraints.

The decision should be based on comparisons of:

• structural behaviour • economic aspects • aesthetics.

In comparing costs, both initial costs and costs associated with maintenance during the life of the structure should be considered. The time required for execution, which in steel bridges is generally shorter than in prestressed concrete bridges, may also influence the decision.

In the past, concrete bridges could not compete with steel bridges for medium and long spans due to the lower efficiency (strength/dead load) of concrete solutions. With the development of prestressed concrete it is not a straightforward decision to decide between a concrete and a steel solution for medium span (about 40 to 100m) bridges. Even for long spans between 200 and 400m, where cable stayed solutions are generally proposed, the choice between a concrete, steel or composite bridge superstructure is not an easy task.

The choice between a steel and a concrete solution is sometimes reconsidered following the contractors' bids to undertake the bridge works.

Generally speaking, steel solutions may have the following advantages when compared to concrete solutions:

• reduced dead loads • more economic foundations • simpler erection procedures • shorter execution time.

A disadvantage of steel when compared to concrete is the maintenance cost for the prevention of corrosion. However it is now recognised that concrete bridges also have problems relating to maintenance, i.e. relating to the effects of the corrosion of steel reinforcement on the durability of the structure.

Although maintenance costs and aesthetics play a significant role in the design decision, the initial cost of the structure is generally the most decisive parameter for selecting a steel or a concrete bridge solution. Solutions of both types are generally considered, at least at a preliminary design stage.

In Figure 1 the principal components of a bridge structure are shown.

The two basic parts are:

• the substructure • the superstructure.

The former includes the piers, the abutments and the foundations.

The latter consists of the deck structure itself, which support the direct loads due to traffic and all the other permanent and variable leads to which the structure is subjected.

The connection between the substructure and the superstructure is usually made through bearings. However, rigid connections between the piers (and sometimes the abutments) may be adopted, particularly in frame bridges with tall (flexible) piers.

2. THE SUBSTRUCTURE

Piers may be made of steel or concrete. Even in steel and composite bridges, reinforced concrete piers are very often adopted. In some cases, e.g. very tall piers or those made by precast concrete segments, prestressed concrete may be adopted. Piers are of two basic types:

• columns piers • wall piers.

Concrete column piers may have a solid cross-section, or a box section may be the shape chosen for the cross-section (Figure 2) for structural and aesthetic reasons.

Wall piers are generally less economical and less pleasing from an aesthetic point of view. They are very often adopted in cases where particular conditions exist, e.g. piers in rivers with significant hydrodynamic actions or in bridges with tall piers where box sections are adopted.

Piers may be of constant cross-section or variable cross-section. The former solution is usually adopted in short or medium piers and the latter in tall piers where at least one of the cross-section dimensions varies along the length of the pier.

The abutments establish the connection between the bridge superstructure and the embankments. They are designed to support the loads due to the superstructure which are transmitted through the bearings and to the pressures of the soil contained by the abutment.

The abutments must include expansion joints, to accommodate the displacements of the deck, i.e. the longitudinal shortening and expansion movements of the deck due to temperature.

Two basic types of abutments may be considered:

• wall (counterfort) abutments • open abutments.

Counterfort wall abutments (Figure 3 and 4) are adopted only when the topographic conditions and the shapes of the earthfill are such that an open abutment (Figure 5) cannot be used. They are generally adopted when the required height of the front wall is above 5,0 to 8,0m (Figure 4). If the depth is below this order of magnitude, counterfort walls may not be necessary and a simple wall cantilevering from the foundation may be adopted.

The connection between the abutments and the earthfill may include a transition slab (Figure 4) which ensures a smooth surface of the pavement even after settlement of the adjacent earthfill.

3. INTRODUCTION TO THE SUPERSTRUCTURE

It is common in bridge terminology to distinguish between:

• the longitudinal structural system • the transverse structural system.

It should be understood that bridge structures are basically three-dimensional systems which are only split into these two basic systems for the sake of understanding their behaviour and simplifying structural analysis.

The longitudinal structural system of a bridge may be one of the following types which are illustrated in Figure 6:

• beam bridges • frame bridges • arch bridges • cable stayed bridges • suspension bridges.

The types of girder incorporated in all these types of bridges may either be continuous i.e. rolled sections, plate girders or box girders, or discontinuous i.e. trusses.

Beam bridges are the most common and the simplest type of bridge (Figure 6a), whether they use statically determinate beams (simply supported or Gerber beams) or continuous beams. Simply supported beams are usually adopted only for very small spans (up to 25m). Continuous beams are one of the most common types of bridge. Spans may vary from small (10 - 20m) to medium (20 - 50m) or large spans (>

100m). In medium and large spans continuous beams with variable depth section are very often adopted for reasons of structural behaviour, economy and aesthetics (Figure 1).

Frame bridges are one of the possible alternatives to continuous beams (Figure 6b). Avoiding bearings and providing a good structural system to support horizontal longitudinal actions, e.g. earthquakes, frames have been adopted in modern bridge technology in prestressed concrete bridges or in steel and composite bridges. Frames may be adopted with vertical piers (the most common type) or with inclined struts (Figure 6b).

Arches have played an important role in the history of bridges. Several outstanding examples have been built ranging from masonry arches built by the Romans to modern prestressed concrete or steel arches with spans reaching the order of 300m.

The arch may work from below the deck, from above the deck or be intermediate to the deck level (Figure 6c). The most convenient solution is basically dependent on the topography of the bridge site. In rocky gorges and good geotechnical conditions for the springings, an arch bridge of the type represented in Figure 6(c) is usually an appropriate solution both from the structural and aesthetic point of view.

Arches work basically as a structure under compressive stress. The shape is chosen in order to minimise bending moments under permanent loads. The resultant force of the normal stresses at each cross-section, must remain within the central core of the cross-section in order to avoid tensile stresses in the arch. Arches are ideal structures to build in materials which are strong in compression but weak in tension, e.g. concrete.

The ideal "inverted arch" in its simplest form is a cable. Cables are adopted as principal structural elements in suspension bridges where the main cable supports permanent and imposed loads on the deck (Figure 6(e)).

Good support conditions are required to resist the anchorage forces of the cable. In the last few years, a simpler form of cable bridges has been used - the cable stayed bridge.

Cable stayed bridges (Figure 6(d)) have been used for a range of spans, generally between 100m and 500m, where the suspension bridge is not an economical solution. The range of spans for cable stayed bridges is quite different from the usual range of spans for suspension bridges - from 500m to 1500m. Cable stayed bridges may be used with a deck made in concrete or in steel. Generally, cable stayed bridges are designed with very slender decks which are "continuously" supported by the stays which are made of a number of strands of high strength steel.

Three main types of transverse structural system may be considered:

• slab • beam-slab (slab with cross-girders) • box girders for longitudinal structural system which contribute to the transverse structural system.

Slab cross-sections are only adopted for small spans, generally below 25m, or where multiple girders are used for the longitudinal structural system, at spacings of 3 - 4,5m. Beam-slab cross-sections (Figure 1) are generally adopted for medium spans below 80m where only two longitudinal girders are provided. For large spans (> 100m), and also for some medium spans (40 - 80m), box girder sections are very convenient solutions leading to good structural behaviour and aesthetically pleasing bridge structures. Box girders are used in prestressed concrete or in steel or composite bridges.

4. STEEL BRIDGES

4.1 General Aspects

During the industrial revolution of 19th century steel products became more competitive and structural steel began to be adopted for bridge construction. From then on, large truss bridges and suspension bridges where developed. Unfortunately this development was accompanied by several accidents, e.g. the railway bridge over the Tay [1] in 1879 and the Quebec bridge in 1907. The former was rebuilt (1890) with spans of 521m; the Quebec bridge was only rebuilt in 1917.

Truss girders or arches built by truss systems have been widely adopted. An example of an arch-truss bridge designed by G Eiffel (the designer of the famous Paris tower) is presented in Figure 7. This bridge, built in 1868 in Oporto over the Douro River, Portugal, has a central span of 160m.

It is interesting to note that one of the commonest types of modern steel bridge - the box girder bridge was first introduced in bridge engineering in 1846 by Stephenson with the "Britannia Bridge" (a cast iron 142m span box girder bridge), yet was only fully developed after the Second World War. The knowledge of aeronautical engineering of thin-walled structures was used. Between 1969 and 1971 several accidents occurred to box girder bridges, e.g. Vienna bridge over the Danube (1969), Milford Haven bridge in the United Kingdom (1970), Melbourne bridge in Australia (1970) and Coblenz bridge in Germany (1971). As a result a large research effort was made over the last two decades to investigate the basic structural element of these bridges - the stiffened plate. The behaviour of stiffened plates is now sufficiently known for safe large box girder bridges to be designed in steel. Special consideration during erection and execution phases is given to all aspects of structural stability.

Three basic types of structural elements are adopted for steel bridge superstructures:

• Beam and Plate Girders • Truss Girders • Box Girders.

Plate girder bridges with only two girders, even for very wide decks (Figure 8), are very often preferred for the sake of simplicity [2]. However, in bridge construction, a classical solution consists in adopting several I beams (hot rolled sections for small spans - up to 25m) with 3,0 to 4,5m spacing. Diaphragms may be provided between the beams (transverse beams) to contribute to transverse load distribution and

also to lateral bracing. The top flanges of the beams have continuous lateral support against buckling provided by the deck.

4.2 Deck Systems

There are two basic solutions for the deck [3] - a reinforced concrete or partially prestressed concrete slab and an orthotropic steel plate (Figure 9). In the former the slab may act independently of the girders (a very uneconomic solution for medium and large spans) or it may work together with the girders (composite bridge deck). The composite action requires the shear flow between the slab and the girders to be taken by shear connectors.

Concrete decks are usually more economic than orthotropic steel plates. The latter are only adopted when deck weight is an important component of loading, i.e. for long span and moveable bridges.

The orthotropic plate deck, acting as the top flange of the main girders, gives a very efficient section in bending. The deck is basically a steel plate overlain with a wearing surface which may be concrete or mastic asphalt. The steel plate is longitudinally stiffened by ribs which may be of open or closed section. Transversally, the ribs are connected through the transverse beams (Figure 9) yielding a complex grillage system where the main girders, the steel plate, the ribs and the floor beams act together.

Top flanges of box girders, e.g. in Niteroi bridge (Figure 10) with a 300m span [4] (the largest in the world for a box girder bridge) or in the deck of cable stayed bridges (Figure 11) [5] or suspension bridges like the Humber bridge (Figure 12) with a lightweight wearing surface give a deck of very low dead load which makes this type of solution very suitable for long spans [4,8]. The biggest disadvantage of orthotropic steel plate decks is their initial cost and the maintenance required when compared to a simple concrete slab. However, for box girders the maintenance cost may be lower than for an open orthotropic deck.

5. PLATE GIRDER BRIDGES

Plate girder bridges can provide a very competitive solution for short and medium span bridges. They are almost always designed to act compositely with the concrete slab.

The plate girders are fabricated with two flanges welded to a thin web which usually has transverse stiffening and may have longitudinal stiffening.

Three types fo bridge cross-section may be used. For shorter spans, up to 60m, multiple girders at spacings of 3 to 4,5 m enable a simple reinforced concrete slab to be used, as shown in Figure 13(a).

For medium spans (50 to 100 m) it is usually more economic to use only two plate girders, Figure 13(b). A prestressed concrete slab, usually of varying depth, may be used that sits directly on the two girders.

Alternatively cross girders may be adopted with twin longitudinal girders that support the slabs at 3 to 4,5 m centres.

The complexity of fabrication of the plate girder is primarily controlled by the web slenderness (depth/thickness ratio). For short spans a low slenderness is feasible with a web that is unstiffened except at cross bearing positions and supports. For medium spans the web will usually need to be of intermediate slenderness and require vertical (transverse) stiffening. For larger spans the web is likely to require both transverse and longitudinal stiffening, as shown in Figure 13(b).

The distance between transverse stiffeners is of the order of magnitude of the depth of the girder. Where they are required, 1 to 3 longitudinal stiffeners are usually provided.

In sections at supports, it is essential to adopt vertical stiffeners to resist the high reaction forces.

One of the basic requirements when designing plate girder bridges is the bracing system (Figure 13b and 14), which is required for all but the simplest structures. The bracing:

• provides lateral stability to the girders, particularly during erection • supports the horizontal shear forces due to horizontal actions (wind, earthquakes) • works as a transverse load distribution system. • takes part in the shear flows due to torsion from eccentric loading or plan curvature.

The bracing system generally includes:

• horizontal lateral bracing • intermediate cross frames - diaphragms.

The former (Figure 14) consists of a set of crossing diagonal members and is located near the bottom or near the top and the bottom flanges; the bridge deck may act as a horizontal bracing.

The latter are a set of bracings (trusses) normal to the bridge axis - Figure 13 - which provide resistance to the deformation of the overall cross-sections of the bridge.

In modern bridge construction several simplifications have been tried in order to reduce, as much as possible, the complexity of the bracing systems. In some cases, the horizontal bracing system located near the bottom flanges has been eliminated. The ultimate simplification consists in avoiding the intermediate cross frames completely. This is only possible if the lateral stability of the girders is guaranteed and the horizontal forces are taken by other elements of the superstructure.

6. TRUSS GIRDER BRIDGES

A truss girder may be adopted in some cases as an alternative to a plate girder. Although less commonly used in modern construction because of their high fabrication content, they may still be an economic solution for large spans, say between 100 and 200 metres.

A plane truss girder may be considered as a deep beam where the flanges are the compression and tension chords of the truss and the web of the beam is replaced by an open triangular system which resists the shear forces.

Several types of truss girders are used in bridge design. Some typical examples are shown in Figure 15. Truss girders may be adopted in simply supported (Figure 15) or continuous spans (Figure 16 and 17).

Bracing systems are required in truss girder bridges, since truss girders can only resist forces in their planes.

Truss girders working from above the deck (Figure 16) have been extensively used in railway bridges, even for medium spans of the order of 40 to 100 metres [6].

From an aesthetic point of view, it is important to reduce as far as possible the number of bar elements in the truss girder. If possible the simplest triangular system (Warren type) yields the best appearance when the bridge is viewed from skew angles (Figure 16).

Truss chords and diagonals are made using hot rolled sections generally of an open shape for simplicity of connections. However, tubular cross-sections may be adopted, for example, for the chords.

An example is shown in Figure 17 which represents the bridge over the river Fulda in Germany, near Kassel [7]. In this bridge, a Warren type truss was used with a maximum span/depth ratio of 23,8. The deck is an orthotropic plate giving a reduced dead weight for the superstructure.

7. BOX GIRDER BRIDGES

For long spans (say, in excess of 100m) box girders are, in general, the most common and efficient type of bridge superstructure. Built with an orthotropic plate deck to reduce the dead weight of the bridge, or with a concrete slab to obtain a composite cross-section, box girders have many structural advantages when compared to plate girders and truss girders. Some of the advantages are:

• high torsional rigidities • wide top and bottom flanges to carry longitudinal forces • large internal space to accommodate services • simple maintenance due to easy access to the interior of the superstructure • better appearance due to high slenderness and smooth bottom surfaces.

Due to the high torsional rigidity of this type of cross-section, box girders are a very convenient solution for bridges curved in plan.

For large spans, the depth of continuous box girder bridges may vary along the span giving improved structural efficiency to accommodate the large bending moment at the supports (Figure 10).

The cross-section may consist of a single cell box, with vertical or inclined webs, or of a multiple cell box (Figure 18). Other possibilities consist of using, for example, a single cell with inclined struts to support large overhangs (Figure 19).

For medium spans, a type of box girder deck very common in bridge construction, e.g. in North America, is the composite box girder deck made of several parallel boxes interconnected by a reinforced concrete slab deck (Figure 20). Composite action between the box girders and the reinforced concrete slab is obtained through shear connectors.

The two flanges associated with each web in composite box girder bridges may be quite narrow because they only need to transfer the load to the web and to accommodate the shear connectors. A minimum flange width may therefore be defined by edge distances and clearances for automatic welding of shear connectors.

Load bearing diaphragms are necessary at supports to transfer the reaction forces. In addition, even in small box girders, it is good practice to adopt intermediate cross frames (say, at 10m to 15m apart) to avoid distortion of the cross-section under eccentric loading (Figure 21). It should be noted that during construction some "box" girders have open sections and so will be subjected to distortion under eccentric loading. Figure 22 summarises the distortions that can occur in open topped boxes during construction. A

top bracing between the top flanges and/or a cross diagonal bracing between the webs is generally convenient to overcome the distortion effects during execution. The diagonal bracing may consist of small size angles welded to plate stiffeners.

The use of composite box girders in wide bridges with long spans is possible with single cell boxes. Internal cross trusses may be used, not only to maintain the shape of the cross-section (avoiding distortion) but also to support longitudinal stringers for the reinforced concrete slab. A solution of this type is shown in Figure 23 [7].

For long spans an orthotropic plate deck is preferred to reduce the dead load of box girder bridges. A solution with a rectangular box girder bridge with a main span of 200m is given in Figure 24 which shows the "Europe Bridge" in Austria [7].

The use of box girders is not restricted to beam bridges. Slender box girders in cable stayed bridges have been used with orthotropic plate decks (Figure 11). Although, in the last few years, concrete box girder decks have been shown to be an economic solution for some cable stayed bridges, steel box girders are the most convenient solution for long spans. Compared to open sections, box girder decks in cable stayed bridges present a significant advantage in respect of aerodynamic stability. The advantage is associated with a higher natural frequency of torsional vibration of the deck avoiding an interaction with the fundamental mode corresponding to vertical vibrations (flexure mode). The risk of flutter instabilities is thus eliminated.

For reasons similar to those given for cable stayed bridges, slender steel box girders with orthotropic plate decks have been adopted in modern suspension bridges.

The cross-section of the Humber bridge, where a steel box girder was adopted weighing only 2,6 kN/m2 is shown in Figure 12. The same aerodynamic advantages pointed out for box girder decks of cable stayed bridges, are valid for suspension bridges.


• A wide range of factors should be considered when deciding on the type of bridge for use at a particular location, e.g. spans, execution processes, local conditions, foundation constraints.

• Steel bridges generally have the following advantages: reduced dead loads, economic foundations, simple erection procedures, short execution times.

• The basic parts of a bridge are the superstructure consisting of the deck structure and the sub-structure consisting of the piers, abutments and foundations.

• The longitudinal system of a bridge may be one of the following types: beam, frame, arch, cable stayed or suspension.

• There are three main types of bridge transverse systems, slab, beam-slab or box girder. • Bridge superstructures may use the beam and plate girder, truss girder or box girder structural

systems. • Deck systems use a reinforced concrete slab, with or without cross-girders, or a partially

prestressed concrete slab, or an orthotropic steel plate.

9. REFERENCES

[1] Wittfoht, H., Triumph der Spannweiten, (Spanish ed. - Puentes - Ejemplos Internacionales) Ed. Gustavo Gili, Barcelona, 1975.

[2] Vevey, Bulletin Technique, 1978.

[3] Alvarez, R., La estructura metalica hoy, Libreria Tecnica Bellisco, 1975.

[4] Pfeil, W., "Pontes" Ed. Campus Ltd, Rio de Janeiro, 1983.

[5] Walther, R., Ponts Haubanés, Presses Polytechniques Romandes, 1985.

[6] Reis, A. and Abecasis, T., Railway Bridge over the River Zezere, preliminary Design Report, Grid Consulting Engineers, 1990.

[7] O'Connor, C., Design of Bridge Superstructures, John Wiley & Sons, 1971.

[8] Gimsing, Niels, Cable Supported Bridges, John Wiley & Sons, 1983.

Note: A general list of wider reading is given at the end of Lecture 1B.6.2.


OBJECTIVE/SCOPE

To continue the introduction to steel and composite bridges. To describe footbridges, moving bridges and service bridges. To provide initial guidance in selection of bridge form and span.

PRE-REQUISITES


RELATED LECTURES

Lectures 15B: Structural Systems: Bridges

SUMMARY

This lecture continues the introduction to steel and composite bridges started in Lecture 1B.6.1. It describes three types of special bridge, highlighting some of the features in design. Footbridges are narrow, lightly loaded structures frequently in visually sensitive locations. Moving bridges are subject to particular constraints of geometry and mass. Services bridges offer special opportunities for innovative design. The lecture concludes with some guidance on the appropriate selection of bridge form and on the determination of optimum spans for viaducts.

1. INTRODUCTION

In Lecture 1B.6.1 attention was concentrated both on the principal design parameters and the various structural forms that a designer may consider when carrying out the preliminary or conceptual design of a bridge. It is probably safe to say that the large majority of bridges are fixed structures carrying a road or railway and it is easy to see how the types discussed can be used for such bridges. In this second part of the Lectures 1B.6, attention will first be given to particular considerations affecting some special types of bridge. Three types will be discussed - footbridges, moving bridges and service bridges (pipelines, etc.). Some guidance on choice of bridge type and span is also provided.

2. FOOTBRIDGES

Footbridges are needed where a separate pathway has to be provided for people to cross traffic flows or some physical obstacle, such as a river. The loads they carry are, in relation to highway or railway bridges, quite modest, and in most circumstances a fairly light structure is called for. They are, however, frequently required to give a long clear span, and stiffness then becomes an important consideration. These bridges are often very clearly on view to the public and then the appearance merits careful attention. Steel offers economic and attractive forms of construction which suit all the requirements demanded of a footbridge. Figure 1 gives schematic views of a range of structural forms in steel.

Like any other bridge, footbridges must be long enough to clear the obstacle which is to be crossed and high enough not to interfere with whatever passes beneath the bridge. However, the access route onto the footbridge is often quite different from that which is familiar to the designer of a highway bridge: there is no necessity for a gentle horizontal alignment (indeed the preferred route may be sharply at right angles to the span). Structural continuity is, therefore, less common; the principal span is often a simply supported one.

Provision of suitable access for wheelchairs and cyclists is often specified for footbridges, see Figure 2. Access ramps must be provided and restricted to a maximum gradient. The consequent length of ramps where access is from the level of the road over which the bridge spans is generally much longer than the bridge itself. The form of construction suitable for the ramps may have a dominant influence on the form of the bridge.

As shown in Figure 3, the width of a footbridge is usually quite modest, just sufficient to permit free passage in both directions for pedestrians. Occasionally the bridge will have segregated provision for pedestrians and cyclists, in which case it will need to be wider.

Parapets are provided for the safety of both the pedestrians and traffic below. Footbridges over railway lines may be required to have higher parapets and be provided with solid panels directly over the rail tracks.

3. MOVING BRIDGES

3.1 General

A fixed bridge to take a highway or railway over a navigable waterway in flat country will require very long approach works. Even a typical highway with a limiting approach gradient of 4%, will require an approach length of 750m each side to give 30 metres clearance (not an unusual figure for a waterway which can be navigated by seagoing ships). This would be increased by the construction depth of the bridge. The length for a railway bridge, with its shallower gradients, would be even longer.

An alternative is to keep the bridge at low level and design it to open to allow the passage of ships. The primary advantage is that the construction cost of an opening bridge is almost invariably much less than that of a high level bridge (and very much less than that of the other possibility, a tunnel). In addition, in flat country for which the comparison is being made, a high level bridge can be visually very intrusive. The main disadvantage of a moving bridge, of course, is the delay to traffic when a bridge is opened to shipping; further disadvantages include the need for manning and maintenance of the opening system, the risk of failure thus disrupting either shipping or surface traffic, and the risk of ships colliding with and damaging the structure.

When the highway and waterway are main routes, there may be no alternative to using either a high level bridge or a tunnel. However, where an element of delay is acceptable, moving bridges are commonly used. The Netherlands and the flat eastern counties of the UK are two regions where there are many such structures.

The design of moving bridges is a highly specialised subject and can only be covered very briefly in general terms in this lecture. Modern moving bridges are likely to be one of three types - bascule, swing or lift, with bascule bridges probably being the most common. The main features of each are discussed briefly below.

3.2 Bascule Bridges

A bascule bridge consists of one or two cantilever arms (or "leaves") which either pivot about horizontal axes at abutment piers (Figure 4a, b and d) or roll backwards on a track (Figure 4c). Normally, such bridges on railways or major highways are of single leaf construction, since they then behave as simply supported girders for carrying traffic loads; if a double leaf configuration is used, even the traffic loads are carried on cantilever structures which, in consequence, have to be of much stronger construction.

The main advantage of a bascule bridge is its efficiency in operation. Bascule leaves are fast to raise and lower, and for the passage of small vessels, need only be raised to part height, thus speeding the operation still further. Furthermore, unlike the swing bridge (see Section 3.3) they operate within the shadow area of the structure.

One disadvantage of the simple pivoting bascule bridge, as shown in Figures 4 (a) and (b), is that the mass of the cantilever has to be balanced whilst pivoting in order to keep the power requirements to a reasonable level,. This means either a significant backspan, with a very deep abutment pier to accommodate it after raising, or the use of a very substantial counterweight. This problem is partly overcome with the "Dutch drawbridge" type (Figure 4(d)) in which the counterweight is mounted on an overhead structure and thus does not need deep abutments to accommodate it.

A further disadvantage is the high power requirement for operating in adverse weather conditions. High winds blowing across a river cause very large forces on the bascule leaves, and snow loading will of course increase the raised mass without any compensating counterweight; additional demands arise on the drive and braking systems. It must be pointed out, of course, that the high reserve of power required for these conditions contributes materially to the efficient operation under normal conditions.

Whilst structurally efficient in most locations, a wide bascule bridge can give problems for a highly skewed crossing since the non-symmetrical shape of the leaves results in unbalanced forces during raising.

3.3 Swing Bridges

A swing bridge pivots about a vertical axis until the superstructure is aligned clear of shipping lanes (Figure 5). The main advantage of a swing bridge is that it probably has the lowest power requirements of any form of moving bridge. If the bridge is symmetrical (equal length swinging arms), wind effects during swinging are small since they will be largely balanced on the two arms and snow loading does not require additional power - the increased inertia merely results in a slightly longer time for swinging due to slower acceleration and braking.

Structurally, a swing bridge is efficient, in that it can be made as a simple girder (either plated or truss) which cantilevers each side of the pivot pier during swinging, but rests on the abutments to give a two-span continuous beam when carrying traffic; if the cantilever span during swinging is excessive it is comparatively easy to design it as cable-stayed with a tower above the pivot pier. The primary purpose of the stays is to carry the dead load during swinging. A skew crossing is no problem for a swing bridge - indeed, it can even be an advantage since it reduces the arc of swing.

The main disadvantages of a swing bridge are the comparatively long time required for swinging from the traffic to the shipping position, and the large plan area required to accommodate the structure when opened for shipping; once swung, of course, it has the advantage that the vertical clearance ("air draught") is unlimited. A swing bridge will normally have to be fully swung for any vessel, regardless of size, since a partial opening will result in the risk of the vessel striking the structure. In addition, since there is usually only one shipping channel, the "backspan" is operationally unnecessary, and the additional structure is expensive. The extra structural cost can be minimised by making the backspan shorter and counterweighting it, but this will reduce some of the advantages listed above - for example, the wind load will no longer be balanced and hence more power will be needed to drive the bridge in high winds. Furthermore, snow loading will put an unbalanced vertical load on the structure. Perhaps the "ideal"

location for a swing bridge is in a river with an island exactly in the middle, and shipping lanes either side!

3.4 Lift Bridges

In lift bridges the span is lifted up towers at each end to clear a shipping channel (Figure 6). Structurally, lift bridges are very efficient, since they are simply supported spans both in-service and during the raising operation. They are designed as girders (either plated or truss) and since they do not have to operate as cantilevers in any condition they can provide much longer opening spans than bascule or swing bridges.

The main disadvantage of a lift bridge is that it can only give as much vertical clearance as is provided by the height of the towers; hence, if very large clearances are required, the towers become very expensive. As for bascule bridges, however, there is no need to raise the span fully for small vessels, thus improving the efficiency of operation. Furthermore, although the dead weight of the lift span will, of course, be counterweighted, an allowance in the drive system has to be made for the possibility of snow on the deck during raising.

3.5 Other Types of Moving Bridge

Other types of moving bridge which have been used in the past (and occasionally at present) are:

• Floating bridges, in which a section can be floated out on pontoons to enable the passage of vessels.

• Transporter bridges in which a car carrying vehicles is suspended from a trolley on, and traverses, a beam at a high enough level to clear shipping.

• Sliding bridges, in which the bridge structure is slid back from the river in line with itself.

All these types have serious operational and other disadvantages for almost all present-day applications.

4. SERVICE BRIDGES

Many, if not most bridges, whether road, rail or footbridges, also carry at least some public utility services (electricity, telephone, water, gas, etc.). Provision for carrying these services varies with the type of bridge - for instance, box girders provide an obvious area for routing them (although care must be taken to provide for accidents - a flooded box girder arising from a fractured internal water main could be disastrous!) On plate girder bridges it may prove possible to carry the services within the footpath, or hanging from cross-girders if the bridge is of that form of construction.

In this section, however, attention is focused on pure service bridges, whose purpose is only to carry a utility. Clearly, a service bridge may be of any of the fixed bridge types already described, but there are certain special considerations. The loading is usually very light compared with road or rail traffic, and hence some of the problems of footbridges arise here also. Perhaps the commonest form of service bridge is a simple light truss, although aesthetic considerations can rule this out in certain locations.

An interesting variation of the simple truss has been employed on occasion for services which require pipelines, e.g. gas or water. This involves making the truss chords from tubular members which service as the actual pipelines - one example of this is a tubular space truss of triangular cross-section carrying high pressure gas in its chords (Figure 7). A small problem occurred in that the tubular chords gave rise to a low level of aerodynamic excitation; whilst this was no immediate problem, there was concern over long term fatigue effects. A simple aerodynamic change to the section was devised to eliminate the excitation.

The use of cable structures is also common for service bridges. Sometimes the service itself provides this - an overhead electricity transmission line is, in effect, a large series of spans using the conductors as bridges. On long electricity crossings, it may be necessary to take special measures to remove any tendency to aerodynamic oscillation. Cable structures can also be sued for pipeline ridges, where the length is too great to allow the pipe to span unaided. In such a case, both suspension and cable-stayed bridges have been used, sometimes using the pipeline itself as the stiffening girder, and sometimes providing a separate girder.

It is likely, however, that if a suspension bridge type is chosen the dead weight will be so low that the structure will be unacceptably flexible. Furthermore, it will be very weak in the transverse direction when subjected to wind loading. A simple way of correcting both faults is to introduce two further cables on either side of the pipeline (or separate girder if provided), inclined downwards from it, and tensioned against the main suspension cable (Figure 8). This form of structure is very light, and well suited for use in areas where access is difficult for transporting heavy pieces.

5. GUIDANCE ON INITIAL DESIGN

5.1 Selection of Bridge Form

Each form of bridge is suited to a particular range of spans, see Figure 9, which also records the longest span for each type of construction.

Suspension or cable stayed bridges are the only forms capable of achieving the longest spans. They are clearly less suitable for road or rail bridges of short or medium span. However, they can be appropriate for shorter span footbridges, partly because they do not have any concentrated loading that requires an expensive stiffening girder and partly for aesthetic considerations. (It should be noted that the same special consideration which is needed for long spans, such as aerodynamic stability, needs to be applied to steel footbridges).

Suspension bridges are still used for the longest spans where intermediate piers are not feasible. The cables are subjected to very high tension and are tied to the ground, usually by gravity foundations sometimes combined with rock anchors. Thus ground conditions with rock at or close to the surface of the ground are essential.

Cable stayed bridges are of suspension form with normally straight cables which are directly connected to the deck.

The structure is self anchoring and, therefore, less dependent on good ground conditions. However, the deck must be designed for the significant axial forces from the horizontal component of the cable force. The construction process is quicker than for a suspension bridge because the cables and the deck are erected at the same time.

Bridge types, such as arches or portals, may be suitable for special locations. For example, an arch is the logical solution for a medium span across a steep-sided ravine. A tied arch is a suitable solution for a single span where construction depth is limited and the presence of curved highway geometry or some other obstruction conflicts with the back stays of a cable stayed bridge.

Portal frame bridges are usually suitable for short or medium spans. In a three span form with sloping legs, they can provide an economic solution by reducing the main span; they also have an attractive appearance. The risk of shipping collision must be considered if sloping legs are used over navigable rivers.

Cantilever trusses were used during the early evolution of steel bridges. They are rarely adopted for modern construction.

Haunched girders are frequently used for continuous structures where the main span exceeds 50m. They are more attractive in appearance and the greater efficiency of the varying depth of construction usually more than offsets the extra fabrication costs.

Flat girders, i.e. girders of constant depth, are used for all shorter span bridges of both simple spans and continuous construction up to spans of around 30m. Rolled sections are feasible and usually offer greater economy. Above this span fabricated sections will be required.

Both haunched and flat girders can be either plate girders or box girders. Development in the semi-automatic manufacture of plate girders has markedly improved their relative economy. This form of construction is likely to be the preferred solution for spans up to 60m or so, if depth of construction is not unduly limited. Above 60m span, and significantly below that figure if either depth of construction is limited or there is plan curvature, the box girder is likely to give greater economy.

5.2 Selection of Span

For major crossings, the governing span is likely to be controlled by the local topography. Even for minor crossings the physical size of the obstacle to be crossed will be the biggest determinant of span.

However, for multispan viaducts a range of spans is possible and the engineer should seek to make the most economic choice. The table below summarises the factors which influence this choice.

Factor Reasons

Location of obstacles Pier positions are often dictated by rivers, railway tracks and buried services.

Construction depth Span length may be limited by the maximum available construction depth.

Relative superstructure and substructure costs Poor ground conditions require expensive foundations; economy favours longer spans

Feasibility of constructing intermediate piers in river crossings

(a) Tidal or fast-flowing rivers may preclude intermediate piers

(b) For navigable waterways, accidental ship impact may preclude mid-river piers.

Height of deck above ground Where the height exceeds about 15m, costs of piers are significant, encouraging longer spans

Loading Heavier loadings such as railways encourage shorter spans

Table 1 Factors which influence choice of span for viaducts

For long viaducts it is worthwhile to carry out initial costed designs for different spans to determine the most economical combination of superstructure and substructure costs. The outcome of a typical study is shown in Figure 10. Typical optimum spans are shown below.

Conditions Highway Railway

Simple foundations

(spread footing or short piles)

25-45 20-30

Difficult foundations (piles 20m long)

35-55 25-40

Piers 15m high 45-65 30-45

Table 2 Typical optimum span ranges for viaducts


• Steel footbridges are light and economic structures that offer considerable opportunities for attractive and innovative design.

• In flat countries moving bridges offer considerable economies over fixed bridges or tunnels. • Moving bridges are usually bascule bridges, swing bridges or lift bridges. Each offer particular

advantages and disadvantages and require attention in design to key features. • The limited performance specifications for service bridges can result in exceptional structural

solutions. • Suspension, cable-stayed, arch, portal, cantilever and girder bridges all have preferred span

ranges. Choice of structural form is one of the most important initial design decisions. • Although many spans are dictated by outside constraints, it is possible to optimise the choice of

spans for viaducts.


1. The Steel Construction Institute. The Steel Designers Manual, 5th Edition, Blackwell Scientific Publications, Oxford, 1992.

2. Iles, D C. The Design of Footbridges. British Steel General Steels, London, 1993. 3. Alvarez, R. La Estructura Metálica Hoy, Librería Técnica Belliso, 1975. 4. Mason, J. Pontes Metlicas e Mistas em Viga Recta, Livros Tcnicos e Cientificos, Rio de Janeiro,

1976. 5. Homberg, H., Trenks, K. Drehsteife Krenzwerke, Sprinzer Verlag, 1962. 6. Hambley, E. Bridge Deck Behaviour, London Chapman and Hall, John Wiley & Sons, 1976. 7. Cusens, A., Pama, R. Bridge Deck Analysis, John Wiley & Sons. 8. Badoux, J. Conception des Structures Metaliques, Partie D, Dimensionnement des ponts. ICOM-

Institut de la Constructin Metalliques cole Polytechnique Federale de Lausanne. 9. Johnson, R P. Composite Structures of Steel & Concrete, Volume 2 - Bridges, SCI P-051,

Collins, 1986. 10. 2nd International Symposium on Steel Bridges, Paris, April 1992. 11. The Steel Construction Institute Design Guides for Bridges:

SCI P065 Design Guide for Continuous Composite Bridges: 1 Compact Sections, Iles DC, 1989

SCI P066 Design Guide for Continuous Composite Bridges: 2 Non-Compact Sections, Iles DC, 1990

SCI P084 Design Guide for Simply Supported Composite Bridges, Iles DC, 1991

SCI P204 Replacement Steel Bridges for Motorway Widening (SCI in association with BCSA and British Steel General Steels), Iles DC, 1992

SCI P208 Motorway Widening: Steel Bridges for Wider Highway Layouts, Iles DC, 1993.

12. Brown, CW. Constructional Steel Design: An International Guide, Elsevier Applied Science, London, 1992.

13. Godfrey, G B. Jointless Bridges in Composite Construction, Steel Construction Today, Volume 3 No. 1, Blackwell Scientific Publications, Oxford, 1989.

14. Tatsumi, M. Long Span Steel Bridges in Japan prsented at Pacific Structural Steel Conference, Japanese Society of Steel Construction, 1992.

Lecture 1B.7.1: Introduction to the Design of Multi-Storey Buildings: Part 1

OBJECTIVE/SCOPE

To present preliminary topics in designing multi-storey buildings.

PREREQUISITES

An understanding of design philosophies and structural arrangements.

RELATED LECTURES



Lecture 1B.2.2: Limit State Design Philosophy and Partial Safety Factors


SUMMARY

The lecture gives a brief description of the fundamental components of a building frame. It presents different structural arrangements to resist horizontal and vertical loadings. Finally, consideration is given to the question of fire protection.

1. INTRODUCTION

A multi-storey building must resist the combined effects of horizontal and vertical loads; it is composed of foundations, frameworks and floor slabs.

The framework comprises columns and beams together with horizontal and vertical bracings, which stabilise the building by resisting horizontal actions (wind and seismic loads).

Floor slabs are supported by beams so that their vertical loads are transmitted to the columns. They are made of reinforced concrete or composite slabs using profiled steel sheets. Columns are commonly made of H or hollow hot-rolled steel sections. The use of hollow sections filled with concrete can improve their fire resistance. Beams are commonly made of I and H profiles. Nevertheless, the use of welded built-up sections can offer more rational solutions in some cases.

The usual structural systems belong essentially to two categories: moment resisting frame systems and braced-frame systems, the second being the simplest and, therefore, the most economic solution.

In braced frames, vertical bracings are formed by diagonal members within the steel frame. These bracings may be of different form (cross-braced X shaped; V or inverted V shaped; symmetrical or unsymmetrical portal). Alternatives to steel bracings are the reinforced concrete shear walls or cores.

These main components of multi-storey buildings and their design are described in the following section:

2. THE STRUCTURAL SCHEME

A multi-storey building includes the following structural components (Figure 1):

a. foundations

b. framework

c. floor structures.

Foundations are made of reinforced concrete. The type of foundation is selected according to the features of the ground and the ground conditions.

The framework is the steel skeleton which provides the load-bearing resistance of the structure and supports the secondary elements such as the floor slab and cladding.

All external loads, both vertical and horizontal, are transmitted to the foundations by means of the steel framework. It is mainly composed of vertical elements (columns) and horizontal elements (beams), which may be connected together in different ways. According to the degree of restraint at the beam-to-column connections, the framework can be considered as 'rigid', 'semi-rigid' or 'pin-ended'. For the pin-ended case, the framework must incorporate bracing elements which are located in the rectangular panels bounded by columns and beams.

The floor slabs are required to resist the vertical loads directly acting on them and to transmit these loads to the supporting floor beams. They also transfer the horizontal loads to the points on the framework where the bracing members are located.

The structural arrangement of multi-storey buildings is often inspired by the shape of the building plan, resulting in different solutions (Figure 2). The plan can be rectangular (Figure 2a), L-shaped (Figure 2b), curved (Figure 2c), polygonal (Figure 2d) or perhaps composed of rectangular and triangular elements (Figure 2e).

3. COLUMNS

Columns are the structural components which transmit all vertical loads from the floors to the foundations. The means of transmission of vertical load is related to the particular structural system used for the framework (Figure 4).

The location of columns in plan is governed by the structural lay-out. The most common grid arrangements are square, rectangular, or occasionally triangular, according to the choice of the global structural system (Figure 3). The spacing of columns depends upon the load-bearing resistance of the beams and floor structures. It can vary from 3 to 20m, but is typically in the range of 6 to 10m.

Load transmission from floors to columns may occur directly from the floor beams to the column (Figure 4a), or it can be indirect. Indirect transmission involves the use of major 'transfer' beams (Figure 4b), which resist all the loads transmitted by the columns above.

In suspended systems (Figure 4c), the transmission of vertical loads is much more complicated. It is directly provided by tensile members (ties), hung from the top beam elements which support the total vertical load of all floors. A limited number of large columns provide the transmission of the total load to the foundations.

The choice of location and spacing of columns depends on the structural system which has to harmonize functional and economical requirements.

The shapes of cross-section commonly used for columns can be subdivided into (Figure 5):

• open sections. • hollow sections.

Open sections are basically standard hot-rolled I and H profiles (Figure 5a). Double-T sections can be also built up by welding. Cross-shaped sections can be obtained by welding L profiles, plates or double-T profiles (Figure 5b).

Hollow sections are tubes of circular, square or rectangular cross-section (Figure 5c). They can also be made from plates or double-T profiles by welding (Figure 5d).

Circular and square hollow sections have the advantage that they have the same resistance in the two principal directions, enabling the minimum section dimensions to be obtained. Sometimes hollow sections are filled with concrete, giving an increase in strength and, at the same time, achieving significant fire resistance (> 60 minutes) (Figure 5e). However the beam-to-column connections are more complicated than between I-sections.

4. BEAMS

Beams support the floor elements and transmit their vertical loads to the columns.

In a typical rectangular building frame the beams comprise the horizontal members which span between adjacent columns; secondary beams may also be used to transmit the floor loading to the main (or primary) beams.

In multi-storey buildings the most common section shapes for beams are the hot rolled I (Figure 6a) or H shapes (Figure 6c) with depth ranging from 80 to 600mm. In some cases channels, (either single or double) can also be used (Figure 6b).

Where a greater depth is necessary, built-up sections can be used. Sections fabricated by welding can have double-symmetrical (Figure 6d) or non-symmetrical (Figure 6e) shape, the latter being advantageous for composite steel-concrete sections. By combining plates and/or profiles, box-sections (Figure 6f) or open sections (Figure 6g) can be fabricated.

Sometimes openings in the webs of beams are required in order to permit the passage of horizontal services, such as pipes (for water and gas), cables (for electricity and telephone), ducts (for air conditioning), etc. The openings may be circular (Figure 6h) or square with suitable stiffeners in the web. Another solution to this problem is given by using castellated beams (Figure 6i), which are composed by welding together the two parts of a double-T profile, whose web has been previously cut along a trapezoidal line.

For buildings, the common range for the span to depth ratio is 15 to 30 in order to achieve most efficient design.

In addition to the strength, beams must provide enough stiffness to avoid large deflections which could be incompatible with non-structural components (such as partition walls). For this purpose the maximum mid-span deflection of a beam is usually limited to a fraction of the span equal to 1/400 - 1/500. Where this limitation is too severe, an appropriate initial deformation (camber) equal and opposite to that due to the permanent loads can be pre-formed into the beam.

Steel sections can be partly encased in concrete by filling between the flanges of the section. Partly encased sections are fire resisting without conventional fire protection (Figure 5e). For longer periods of fire resistance, additional reinforcing bars are required.

5. FLOOR STRUCTURES

Floor are required to resist vertical loads directly acting on them. They usually consist of slabs which are supported by the secondary steel beams. The spacing of supporting beams must be compatible with the resistance of the floor slabs. Floor slabs may be made from pre-cast concrete, in-situ concrete or composite slabs using steel decking. A number of options are available:

• conventional in-situ concrete on temporary shuttering (Figure 7a). • thin precast elements (40 - 50mm thick) with an in-situ structural concrete topping (Figure 7b). • thicker precast concrete elements which require no structural topping (Figure 7c). • steel decking acting as permanent shuttering only (Figure 8b). • steel decking with suitable embossments/indentations so that it also acts compositely with the

concrete slab (Figure 8c).

Typical spans for concrete slabs are 4m to 7m, thus avoiding the need for secondary beams. For composite slabs, various cross-section shapes of steel decking are available (Figure 8a). They are classified in three categories according to their load-carrying resistance:

• profiles with a plain trapezoidal shape without stiffeners with a depth up to 80mm (Figure 8c); • profiles with a trapezoidal shape with longitudinal stiffeners both in web and flange with a depth

up to 100mm (Figure 8d); • profiles with both longitudinal and transverse stiffeners with a depth up to 220mm (Figure 8c).

Deck spans range in length from 2 to 4m for the first category, from 3 to 5m for the second category, and from 5 to 7m for the third category. Secondary floor beams can be avoided in the last case.

Permissible spans for steel decking are influenced by conditions of execution, in particular whether temporary propping is used. Such propping is best avoided since the principal advantage of using steel decking, i.e. speed, is otherwise diminished.

To increase the strength and stiffness of the floor beams, a composite steel-concrete system can be obtained by means of appropriate studs welded on the top of the flange (Figure 8f). In this case the slab and beam may be designed compositely using conventional theory.

6. BRACING

Bracing systems are used to resist horizontal forces (wind load, seismic action) and to transmit them to the foundations.

When a horizontal load F (Figure 9a) is concentrated at any point of the facade of the building, it is transmitted to two adjacent floors by means of the cladding elements (Figure 9b).

The effects of loads H acting in the floor slab are distributed to the vertical supporting elements which are located in strategic positions of the structural layout (dotted lines in Figure 8c) by means of an appropriate horizontal resisting element in the floor.

The vertical supporting elements are called vertical bracings; the horizontal resisting element is the horizontal bracing which is located at each floor.

Where horizontal bracings are necessary, they are in the form of diagonal members in the plan of each floor, as shown in Figure 9c).

If steel decking is used, the diagonal bracing can be replaced by diaphragm action of the steel sheeting if it is fixed adequately.

Both horizontal and vertical bracings represent together the global bracing system, which provides the transfer of all horizontal forces to the foundations.

Vertical bracings are characterised by different arrangements of the diagonal members in the steel frame. They are (Figure 10):

a. Single diagonal

b. Cross-braced (X-shaped bracing)

c. Inverted V-shaped bracing

d. Unsymmetrical portal

e. Symmetrical portal

f. V-shaped bracing.

An alternative to steel bracings is provided by reinforced concrete walls or cores which are designed to resist the horizontal forces (Figure 11). In these systems, so-called dual systems, the steel skeleton is subjected to vertical forces only. Reinforced concrete cores are usually located around the stairway and elevator zones.

Normally, the floor slab can be designed to resist in-plane forces to avoid the use of horizontal diagonals. This is the case for in-situ reinforced concrete slabs, or composite slabs with appropriate shear connectors.

7. STRUCTURAL SYSTEMS

To provide resistance to the combined effects of horizontal and vertical loads in a multistorey building, two alternative concepts are possible for the structural system.

The first, so-called 'moment resisting frame system', is a combination of horizontal (beams) and vertical (columns) members which are able to resist axial, bending and shear actions. In this system no bracing elements are necessary. The moment resisting frame behaviour is obtained only if the beam-to-column connections are rigid, leading to a framed structure with a high degree of redundancy. As a consequence of this choice:

• the connections or joints between members are complicated. • the interaction between axial forces and bending moments is critical in column design. • the overall sway deformability of the structure can be too large, as it depends only on the inertia

of the columns.

Typical details of beam-to-column joints for rigid framed systems are shown in Figure 12. They are called 'rigid joints' and their task is to transfer bending moment from the beam to the column. Type (a) can transfer limited bending moments only because the column web can buckle due to local concentration of effects. The presence of horizontal stiffeners in the column web (Type (b)) recreates the cross-section of the beam and the column web panel has to resist the shear force only.

Types (a) and (b) require the execution of welding operations on site. Such operations are not completely reliable and they are also expensive and can cause delay in erection.

A better alternative is to use bolted connections which allow rigid joints to be made without the disadvantages of site welds. Two typical solutions for rigid frame structures, shown in Figures 12c and 12d, are:

• Type (c) is the extended end plate joint. • Type (d) is the cover plate joint.

These solutions allow the most suitable use of connecting methods, i.e. welding in the shop to build up prefabricated elements and bolting in site for connecting them together. This type of joint can be, therefore, called 'shop-welded field-bolted'.

To avoid the practical problems of rigid frame construction, a more convenient solution can be obtained by conceiving the structural behaviour in a different way. The functions of resistance to vertical and horizontal loads are separated in the different 'families' of members, which are grouped in two sub-structures (Figure 13):

a. a simple frame composed by beams pinned together, which is capable of transferring the vertical loads to the foundation (Figure 13a).

b. a cantilever fixed to the ground which resists horizontal forces and transfers their effects to the foundation (Figure 13b).

Sub-structure a. is hyperstatic; beams are bent in the vertical plane, columns are simply compressed, the hinged joints between beams and columns absorb shear forces only.

Sub-structure b. is isostatic; its bracing function can be obtained by means of steel trusses or by reinforced concrete walls. These bracing structures are mainly loaded in shear and bending and their deformability must be checked under serviceability conditions in order to limit sway.

The combination of both sub-structures a. and b. provides the complete structure (Figure 13c), which is able to resist both vertical and horizontal loads.

The main advantages of this solution, the so-called 'braced-frame system', are:

• construction details of joints are very simple, because they act as hinges. • sway deformability of the structure is limited by the bracing system (sub-structure b). • interaction between axial forces and bending moments in the column is virtually absent.

In contrast, some complications arise in the foundation of bracings which must resist the overall horizontal forces with a very small amount of axial compression. High values of eccentricity occur which require large dimensions of the contact area under the foundation.

In these structural systems beam-to-column joints must resist only axial and shear forces. Some typical solutions of joints for pin-ended structures are shown in Figure 14; they are 'shop-welded field-bolted' joints. The most commonly used is the bolted connection between the beam web and the column flange (or web) by means of double angles (Figure 14e, f). They are more economic than the fully welded solutions (Figure 12a, b) for rigid structures and allow simple erection.

8. DESIGN REQUIREMENTS

The design of a structural system for a multi-storey building must to take into account its spatial behaviour.

For the 'braced frame system', which seems to be most convenient for economy and reliability, it is necessary to locate a sufficient number of bracings to allow any horizontal loads however directed to be resisted. For this purpose, the requirements are:

(1) it must be possible to consider any floor system as a plane structure, restrained by the vertical bracings.

(2) bracings, as external restraints of the floor system, must provide a system of at least three degrees of restraint.

(3) the floor system must be capable of resisting the internal forces due to the applied horizontal loads.

To fulfil requirement (1), diagonal bracings must be introduced in the plane of the floor, thus transforming the floor system itself into a horizontal truss.

As an alternative, the slab of prefabricated concrete elements in the floor system can be assumed to resist directly the horizontal forces as a plane plate structure, because its deformability is normally negligible.

Where concrete slabs are used, the erection of the steel skeleton requires particular care, because it is unstable until the floor elements are placed. Temporary bracing is therefore necessary during this phase of execution.

To fulfil requirement (2) the steel truss bracings are active only in their own plane and therefore represent a simple restraint for the floor system. When reinforced concrete bracings are used, they can have one, two or three degrees of restraint, depending upon their resistance to one plane bending (wall), bi-axial bending or bi-axial bending and torsion (core), respectively.

Finally, requirement (3) is fulfilled by evaluating internal forces in the floor elements due to the horizontal loads by considering the location of the vertical bracings.

Figure 15 shows a three-dimensional structure for a multi-storey building with steel bracings. Every point of the floor system is fixed in two directions. In particular, the diagonals connecting points A and B restrain all the points in line '1' in the 'x' direction. The floor bracing is able to receive external forces from both direction 'x' and 'y' and to transmit them to the vertical bracings.

The spatial structure can be reduced to plane sub-structures whose static schemes are shown in Figure 16. The longitudinal facade along row '3' is directly braced in its plane as well as the lateral facades by the transverse bracings of axes 'a' and 'b'.

The pinned joints of the intermediate transverse frames and of the longitudinal frames of axes '1' and '2' are prevented from any horizontal displacement because they are all connected to the vertical bracings by means of the floor bracings. Thus they can be considered as non-sway frames.

Figure 17 represents the spatial structural scheme of a multi-storey building with a reinforced concrete bracing core. It can be considered as an alternative solution of the previous example for the same building, in which the concrete core substitutes both longitudinal and transverse steel bracings.

Two floor bracing systems can be considered:

If all the four walls of the staircase core are structurally effective, the solution of Figure 17a is correct. If only three sides of the staircase core are structurally effective, the transmission of the horizontal forces acting in the longitudinal direction to the longitudinal wall requires the use of additional floor diagonals, as shown in Figure 17b.


• Structural frames for multi-storey buildings consist of an appropriate arrangement of slabs, beams, columns, foundations and bracing.

• There is a wide variety of forms which each of these elements can take to satisfy different detailed requirements.

• Structural arrangements are influenced by the plan shape of the building; the column layout must take account of economic and functional considerations.

• Frames may be moment-resisting (with rigid beam-column connections), but more commonly use simple 'pinned' beam-column connections, lateral stability being provided by an independent bracing system.

• Bracing is required in three orthogonal planes - typically these planes are two non-parallel vertical planes and horizontally within floors, either by the floor slab itself or diagonal bracing.

• Bracing in the vertical plane is most commonly achieved by cross-bracing, or by shear walls for buildings of modest height.


1. Hart, F., Henn, W. and Sontay, H., "Multi-storey Buildings in Steel" Crosby Lockwood Staples, London, 1985.

2. Owens, G. W., Steel Designers' Manual, Blackwell Scientific Publications, Oxford, 1992.

Lecture 1B.7.2: Introduction to the Design of Multi-Storey Buildings: Part 2

OBJECTIVE/SCOPE

To discuss structural arrangements in multi-storey buildings with particular reference to resisting lateral loads.

PREREQUISITES

An understanding of design philosophies, structural arrangements and static and dynamic structural analysis.

RELATED LECTURES


Lecture 1B.2.1 : Design Philosophies

Lecture 1B.2.2 : Limit State Design Philosophy and Partial Safety Factors

Lecture 1B.3 : Background to Loadings

Lectures 1B.4 : Historical Development

Lecture 14.8: Classification of Multi-Storey Frames

Lecture 14.9: Methods of Analysis for Multi-Storey Frames

Lecture 14.10: Simple Braced Non-Sway Multi-Storey Buildings

Lecture 14.14: Methods of Analysis of Rigid Jointed Frames

SUMMARY

This lecture discusses different structural systems (shear frame, shear truss-frame, steel-concrete, tube, etc.). Particular comment is made with regard to ultra high-rise buildings and seismic effects.

1. INTRODUCTION

The use of structural steels in the last century permitted a great increase in the height of building constructions leading to modern high-rise buildings.

For low-rise buildings the most common structural solution is obtained by integrating two different load resisting systems in the same structure:

• semi-rigid or pinned frames, which resist vertical actions only. • steel bracings or concrete walls and cores, which resist horizontal actions.

By means of the so-called 'fourth dimension of steel construction' (which introduces, besides the three geometrical dimensions, the range of available material strength), it is possible to unify the cross-sections of members and, therefore, to obtain optimum and economic solutions for a range of building forms.

For high-rise buildings (up to 120 storeys), different structural systems are used according to the height range:

• up to 30 storeys, concrete wall or core systems. • from 30 to 60 storeys, frame systems. • above 60 storeys, tube systems.

For 'braced' steel frames, different types of bracing can be used according to the structural and functional requirements.

Appropriate calculation models for multi-storey buildings can be used for pin-ended structures and truss bracings.

For seismic resistant steel structures, an excellent performance in terms of strength and ductility can be obtained. The design requirements in such cases correspond to three given limit states: serviceability, resistance to damage, and prevention of collapse.

2. FROM MULTI-STOREY TO HIGH-RISE BUILDINGS

In the last hundred years man has accepted the challenge to increase the size of multi-storey buildings. Height has been increased successfully thanks to the use of structural steels which give suitable mechanical properties in terms of strength and ductility.

The resulting range of buildings extends from multi-storey buildings to tall buildings, and to 'skyscrapers'. The increase in height is gradually changing the skyline of many cities (Figure 1). The development of taller buildings has stimulated the creation of new structural systems, which are more able to provide the increasing resistance needed due to the effects of height. The dynamic action of wind is no longer negligible as the number of storeys increases and becomes as important as the horizontal seismic actions due to earthquakes. Examples of this situation may be found in the high-rise buildings of the United States. In 1965 the John Hancock Center in Chicago was considered the tallest building in the world with 100 storeys and 335m height (Figure 2), excluding the traditional Empire State Building in New York built in 1931 using the structural engineering practice of the time. The innovative structural system of the John Hancock Center consists of a bearing structure around the perimeter which behaves as a framed and diagonal tube.

From 1970, the erection of the twin towers of the World Trade Center in New York commenced which surpassed in height, both the John Hancock Center and the Empire State Building. The twin towers have a square plan and their structural system is called a 'tube in tube', because it is made of an external skin with very close steel columns (framed tube) and an internal core where all vertical facilities are concentrated (stairs, elevators and so on). This concept allowed the building to reach 104 storeys and 411 metres of height (Figure 3).

The supremacy in height of the John Hancock Center was of very short duration. In 1974 the 'Sears Tower' in Chicago became the tallest building in the world, it being 110 storeys and 442m in height (Figure 4a). Its structural system consists of an external framed tube located on the perimeter together with three horizontal trusses, which act as ring belts. A feature of the building is the reduction of its plan area with the height, which transforms the base square into a quasi-rhombic shape, a cross shape and finally a rectangular shape at the top of the building. The variation of the resisting cross-section makes this structure similar to a big cantilever with variable section. It is interesting to observe that the perimeter structure is made of completely prefabricated elements of three spans and two storeys in height which characterise the facade (Figure 4b).

The Sears Tower has now been passed by the Petronas Towers in Kuala Lumpar at 452m and will also be passed in 2001 by the World Financial Centre, Shanghai at 460m. In the last twenty years, many types of multi-storey and high-rise buildings have been erected not only in the USA, but also in Europe and Japan.

3. THE MAIN FEATURES OF LOW -RISE STEEL BUILDINGS

The simplest way to resist both vertical and horizontal loads is to use moment resisting frames (cases 1 and 5 of Figure 5), with floor structures oriented in transverse and longitudinal directions, respectively. This solution is not rational, and therefore not the most economical, because it requires beams and columns with different cross-sections at the various levels. In addition, it is susceptible to too large sway deflections when the number of storeys is greater than 4 or 5.

A better solution is obtained by the use of two different structural systems in the same building (cases 2, 3, 4, 6, 7, 8 of Figure 5), i.e.:-

• semi-rigid or pinned frames, which resist vertical actions only. • steel bracings or concrete walls and cores, which resist horizontal actions.

Both systems are connected together by means of floor structures, which provide a rigid diaphragm at each storey level. The main advantage of this solution is that it makes it possible to unify the shapes of all beams independently of the floor level.

The unification of column sections is also possible, provided that different grades of steel are used (S235, S275, S355) according to the magnitude of stress in the columns. This use of different steel grades is commonly called 'the fourth dimension of steel construction', because it allows, in addition to the three geometrical dimensions, the adjustment of the steel strength in order to optimise the working conditions of the structural members. The unification of the shape of the structural elements is a fundamental pre-requisite for reducing the costs of fabrication and erection.

The first example of the use of the 'fourth dimensions of steel construction' was the IBM Building in Pittsburgh, built in 1965 with three different kinds of steel for the bars of the external lattice bracing, (Figure 6).

4. STRUCTURAL SYSTEMS FOR HIGH-RISE BUILDINGS

The structural system of a high-rise building must resist both gravity and lateral loads, due to phenomena such as wind and earthquake. As the height of the building increases, the lateral loads gradually dominate the structural design.

Figure 7 systematically compares some frequently used steel structural systems on the basis of the structural efficiency, which is measured by the weight of the building [1]. Framed tube structures could be conveniently used in high-rise buildings up to 20 storeys.

Lateral loads due to wind and earthquake produce lateral accelerations. As people normally perceive these accelerations during service conditions, stiffness rather than strength tends to become the dominant factor in buildings of great height. The serviceability limit state can, therefore, be more important than the ultimate limit state.

Four overall groupings of structural systems may be identified (Figure 8). They are:

a. bearing wall system

b. core system

c. frame system

d. tube system.

Each system has different lateral load resisting properties and thus tends to be 'efficient' over a different height range.

The bearing wall system due to the self weight of the structural components (usually concrete), normally becomes inefficient for buildings above 15-30 storeys in height.

The concrete core system has the same disadvantage as the bearing wall system, namely self weight is a limiting factor.

The efficiency of the framed system depends upon the rigidity of the connections and the amount of bracing. Stiffening can be achieved by use of a solid core, shear walls or diagonal bracing. As more bracing is incorporated into the spatial frame, the range of efficient height is increased. The upper limit is in the range of 60 storeys.

The tube system can be thought of as a spatial frame with the vertical elements positioned at the exterior. The range of height efficiency is influenced by the type and the amount of bracing employed in the tube. In general a tube structure is considered the most efficient form for the tallest buildings, i.e. above 60 storeys in height.

From the four basic structural systems, six secondary systems can be derived from a combination of the basic ones (see Figure 8).

The four basic systems are assumed as the prime groups which can be associated to the levels of the structural system hierarchy as proposed by Falconer and Beedle. These primary systems are:-

1. A bearing wall structure is comprised of planar vertical elements, which form all or part of the exterior walls and in many instances the interior walls as well. They resist both vertical and horizontal loads and are mainly made in concrete (see Figure 9).

2. A core structure is comprised of load bearing walls arranged in a closed form where the vertical transportation systems are usually concentrated. This arrangement allows flexibility in the use of the building space outside the core. The core can be designed to resist both vertical and horizontal loads. Figure 10 shows some examples of this system. In the upper part of the figure, there is a central core from which floors are either suspended or cantilevered. In the lower part the cores are separated and connected by the floor structures.

3. A frame structure is usually made of columns, beams and floor slabs arranged to resist both horizontal and vertical loads. The frame is perhaps the most adaptable structural form with regard to material and shape, due to the many ways of combining structural elements in order to give adequate support to the given loading. In the examples of Figure 11, steel frames are combined with concrete walls and cores, or with steel bracings and horizontal trusses.

4. A tube structure is normally characterised by closely spaced exterior structural elements, designed to resist lateral loads as a whole, rather than as separate elements. Alternative schemes could include braced tubes and framed tubes (see Figure 12). Besides the simple tube, tube-in-tube solutions can be also used. These systems allow for more flexibility in the use of interior space, due to the lack of interior columns.

Wall structures as well as cores are usually made of reinforced concrete.

Steel frames can be used together with concrete cores, and/or walls, leading to composite structures, which may be called also 'dual structures'.

When steel frames are braced, different types of bracing can be used according to structural and functional requirements (Figure 13).

The most common are:

• single or double diagonal bracing • vertical or horizontal K-bracing • lattice bracing.

Both K- and single diagonal bracings can be 'eccentric', i.e. the diagonal members do not meet in the nodes.

5. CALCULATION MODELS

5.1 Basic Assumptions

In the design of multi-storey buildings it is usual to refer to a calculation model, which corresponds to an ideal structure having perfect constraints. In contrast, the actual construction details show that the connections between the various members which comprise the structure are considerably different from the assumed idealisations. It is therefore important to point out that any approach to structural design must be based on simplified hypotheses and schemes which make the correlation between the actual structure and model possible. Only the model can be studied by the methods of structural analysis. The results of the analysis will more closely predict the actual behaviour of the structure, the nearer the model represents the structure itself.

A question to ask is whether the introduction of simplifying hypotheses leads to a model whose behaviour is on the safe side or not. It is necessary to check whether the results obtained from the model and, in particular, the ultimate load carrying resistance at collapse, are safe or unsafe.

To answer this question it may be helpful to apply the basic design static theorem. In a structure subject to a set of external forces Fj, αuFj are the values of the loads that, if applied, would produce the collapse of the structure, αu being the actual collapse multiplier. If, for a generic load αFj it is possible to find a distribution of internal forces which balances the external forces, and if the structure also complies everywhere with a given plasticity criterion, then α ≤ αu.

This theorem is valid if the following hypotheses are satisfied:

• Effects of local buckling are absent. • Second order effects have no influence. • Strain values at each point of the structure are lower than those corresponding to material rupture.

A calculation model will be able, therefore, to predict actual behaviour more nearly as the compatibility conditions are more strictly satisfied.

Any solution is however on the safe side, even though compatibility is not complied with, provided that:

• It represents the equilibrium between internal and external forces • It observes the material strength. • The structure has enough ductility, which is necessary to avoid localised fractures, for load values

below those for local or overall structural collapse.

Clearly, once the calculation model has been defined, the stability of members must be checked and, in the case of highly deformable structures, the influence of second order effects on vertical loads must be assessed. Some typical examples of calculation models of steel structures are described below.

5.2 The Pin-Ended Structure

The model of a generic pin-ended structure (Figure 14) can be studied with reference to various positions of the ideal hinges. They can be located, for example, in any one of the three positions shown in Figure 14. Results will be on the safe side provided the dimensions of the various structural elements comply

with the assumed model. From the three cases shown, the following criteria can be deduced for calculating the moments and forces in the columns, beams and connections (sections X-X and Y-Y).

Scheme 1

Columns A and B are simply compressed.

L is the span for calculating the beam moment.

The joint section X-X must resist not only a shear force V = R1, but also a moment M = R1a.

The joint section Y-Y must resist not only a shear force V = R1, but also a moment M = R1 (a + e).

Scheme 2

Column B is compressed (N = R1B + R2B) and subject to a moment M = a (R1B -R2B) concentrated at the central axis.

Column A is compressed (N = R1) and subject to a moment M = R1 a concentrated at the central axis.

L - 2a is the span for calculating the beam moments.

The joint section X-X must resist a shear force V = R1 only.

The joint section Y-Y must resist not only a shear force V = R1, but also a moment M = R1e.

Scheme 3

Column B is compressed (N = R1B + R2B) and subject to a moment M=(R1B -R2B) (a + e) concentrated at the central axis.

Column A is compressed (N = R1) and bent by a moment M = R1 (a + e) concentrated at the central axis.

L - 2 (a + e) is the span for calculating the beam moments.

The joint section X - X must resist a shear force V = R1 and a moment M = R1e.

The joint section Y - Y must resist a shear force V = R1 only.

Each of these three models is on the safe side and can therefore be assumed for calculation. The choice between them is made considering the structural element or the joint which is the weakest part of the structure. The model which minimises the internal forces in that part is chosen, because it is the most safe.

In the first scheme the state of stress in the column is the lowest. It can, therefore, be chosen when columns are oriented according to their weak axis (Figure 15a). Bending effects in the columns are, in fact, eliminated in spite of slight moments in the joints due to a relatively small eccentricity of the bolted connection.

The second scheme is often conservative if the columns are oriented according to the strong axis (Figure 15b). In this case, in fact, eccentricity is greater than half the column depth and it could require an increase in the resistance of the connection. The eccentricity also entails a greater stress in the columns due to bending moments. Their distribution can be evaluated by assuming hinges at the mid-point between floors and by considering the columns fixed by the bracing structure (Figure 16a). Thus, each vertical row can be considered by means of the isostatic scheme shown in Figure 16b. The horizontal reaction Hi is given by rotational equilibrium around the hinge number i:

Hi =

The effects of forces Hi, for each floor and for each column, are resisted by vertical bracings through the floor system. Their intensity is approximately ∆R e/h, wherein ∆R is the difference between the reactions of two beams connected at the column and e/h is the ratio between the hinge eccentricity and the floor height. In the types of structure being considered, as the beam spans are comparable, ∆R depends mainly on any unbalanced accidental loads. Furthermore, as e/h is essentially small, these effects are generally negligible compared with those due to external loads. In contrast, bending moment effects on the columns are not negligible. The corresponding increase in stress must be considered in the calculations.

5.3 The Truss Bracing

Forces acting on bracing structures, such as the effects of wind, earthquakes and geometric imperfections, do not act in a particular direction. Therefore, the scheme of a bracing system has to be designed and calculated for a range of loading conditions.

Referring to the simple truss bracing shown in Figure 17a, the behaviour of a single diagonal system is considered (Figure 17b).

The structure is hyperstatic. Its solution to determine sway displacements is determined by the

compatibility condition (Figure 17c), assuming member as rigid. This assumption imposes the equality ∆AB = ∆CD between the elongation of the tension diagonal AB and the shortening of the compression diagonal CD. If the N-∆ relationship between the axial load N and variation ∆ in the length (Figure 18a) is equal in both tension and compression, then the axial force in both diagonals has the same absolute value. The structure can be considered as the superposition of two isostatic structures working in parallel (Figure 18b) and its solution is straightforward.

The diagonals, however, differ substantially in their behaviour. The compressed bar CD may not have a linear behaviour because, although it remains elastic, it is subjected to buckling and the variation from linear behaviour increases as its slenderness λ increases (Figure 19a). For high slenderness (Figure 19b), the geometric condition ∆AB = ∆CD requires an axial load Nc in the strut which is substantially lower than the axial load Ni in the tie.

There are, therefore, two ways of dealing with bracing. It can be dimensioned so that both diagonals can resist both tension and compression. For this purpose a low slenderness is required (λ ≤ 100), so that the difference in behaviour between tension and compression bars is negligible. This solution is illustrated in Figure 18b: both diagonals cooperate in resisting shear forces. Alternatively, the bracing can be dimensioned by considering the tension diagonal alone. High slenderness is required (λ ≥ 200) in order to ensure that, when the stress reverses and the diagonal becomes a strut, it will remain elastic even if it buckles. Under this condition the bar in compression is redundant and the forces are wholly resisted by the tension bar. Bracings designed in this way are generally more economical, but deformation of the structure is greater. Furthermore, the possibility of buckling of compression diagonals makes this solution inadvisable whenever the bracing is located in the plane of facades or partition walls.

The above considerations are applicable also to other types of bracings.

The bracing shown in Figure 20a, for example, consists of two inclined bars connected to a beam which resists bending. The beam can be calculated by the method indicated in Figure 20b or by that of Figure 20c, according to whether the compression bar is taken into consideration or not. The bracing in Figure 20b, corresponds to members of a truss bearing axial loads only. One diagonal member is one in tension and one in compression. As both members are identical, a check must be made that they can satisfactorily resist the compressive load. In Figure 20c only the tension bar is considered operative. Consequently the beam must also resist bending due to the external force H. In this case also the bracing can be economical, provided the compression bar is sufficiently slender to buckle whilst remaining elastic.

The same approach can be followed for the bracing system shown in Figure 21a. In Figure 21a the bracing bars are designed to act in both tension and compression. This design minimises bending in the beam. Alternatively, in Figure 21c the bracing is designed to take tension only, the member in compression being ignored. This design increases bending in the beam.

6. SEISMIC REQUIREMENTS OF STEEL STRUCTURES

Multi-storey steel buildings are used more and more extensively in regions of high seismic risk because of their excellent performance in terms of strength and ductility. Their performance is due to the mechanical behaviour of materials, structural elements and non-structural components, which is required by the design. The design requirements correspond to the imposition of three given limit states. They are the serviceability, resistance to damage (damageability) and collapse limit states which are included in the new generation of seismic codes, such as the ECCS Recommendations for Steel Structures in Seismic Zones [2] and the Eurocode 8 [3].

The serviceability limit state corresponds to minor frequent earthquakes. It requires that the structure together with the non-structural components should suffer no damage and that discomfort to the inhabitants should be minimal. The first requirement (to avoid structural damage) is fulfilled by designing the structure in elastic range. The second requirement (to avoid non-structural damage and inhabitants' discomfort) is obtained by providing sufficient stiffness to prevent significant deformations.

The "damageability" limit state allows some minor damage to non-structural components due to local large deformations in certain zones. Such damage may occur under less frequent moderate earthquakes.

The collapse limit state is related to severe ground motions due to very infrequent earthquakes. Both structural and non-structural damage is expected, but the safety of the inhabitants has to be guaranteed. The structure must be able to absorb and dissipate large amounts of energy. Different ways can be used to absorb and dissipate energy under very strong ground motions in order to prevent collapse.

7. BEHAVIOUR UNDER HORIZONTAL LOADS

Traditionally, two families of structural systems have been used in multi-storey buildings to resist important horizontal loads (both wind and earthquakes). They are the concentrically braced frames and the moment-resisting frames.

The concentrically braced frame system is widely used both for normal and seismic-resistant steel structures. Vertical cantilever trusses are formed by diagonal bracing elements with coincident centrelines. They resist lateral forces (both winds and horizontal earthquakes) by means of axial forces in the bracing elements, leading to a large stiffness in the elastic range. In these structures the dissipative zones are mainly located in the tensile diagonals, because it is usually assumed that the compression diagonals are buckled.

The inelastic cyclic performance of concentric bracings is rather unsatisfactory due to the repeated buckling of the diagonal members. This buckling produces a progressive reduction of the area of the hysteresis loops, which corresponds to a significant decrease in the capability of the structure to absorb and dissipate energy. This behaviour is illustrated by the shape of the hysteresis loops of a concentric bracing (Figure 22).

Different behaviour arises according to the type of bracing. The types can be classified into three categories: diagonal X-bracings (Figure 23a, b, c), V-bracings (Figure 23d, e, f) and K-bracings (Figure 23g). The X-bracings (Figure 23a) dissipate energy by means of the plastification of both compression and tension diagonals and the degradation is due to out-of-plane buckling, which interacts with local buckling of the cross-section. From this point of view, symmetrical sections (double C, hollow sections) exhibit a better performance than unsymmetrical ones (back-to-back angles). In the V-bracings, the horizontal forces are resisted by both tension and compression diagonals, the last being necessary for equilibrium. From the cyclic loading point of view, only the compression diagonal dissipates energy, whereas the tension diagonal remains elastic.

K-bracings (Figure 23g), on the contrary, cannot be considered as dissipative because the diagonals intersect the column in an intermediate point, thereby including the column in the yielding mechanism.

In summary, for all types of concentrically braced frames, unacceptable large interstorey drifts causing non-structural damage can occur due to the failure of bracings.

The moment-resisting frames have a large number of dissipative zones which are located near to the beam-to-column connections. They resist horizontal forces essentially by bending and energy can be dissipated by means of cyclic bending behaviour.

Beam-to-column connections are usually designed according to the four main types of joints (Figure 24):

Type A, where three plate splices are welded to the column and bolted to the flanges and to the web of the beam.

Type B, where angle splices are bolted both to the column and to the beam.

Type C, with end plate joint with symmetrical extension.

Type D, which is a fully welded joint.

The performance of all types has been found by testing to exhibit sufficient ductility.

Moment resisting frames are widely used for low-rise buildings, but they are generally more expensive than the concentrically braced system for a given height. For medium and high-rise buildings (from 6 to 40 storeys) framed structures exhibit too large elastic deformations under the action of low earthquakes or

wind, producing damage to non-structural elements. Sufficient stiffness can be obtained by adding diagonal bracings to the rigid frame.

From a comparison between the behaviour of concentrically braced and moment-resisting frames, it is concluded that neither of these two traditional systems fulfil contemporary requirements for the three limit states: serviceability, damageability and collapse.

A suitable harmonisation between the lateral rigidity of bracings and the ductility of frames can be obtained using the hybrid framing system of eccentrically braced frames (Figure 25). In this system the horizontal forces are resisted mainly by axially loaded members, but the eccentricity of the layout allows the energy dissipation by means of cyclic bending and shear behaviour in an element known as an active link.

The common type of eccentrically braced frame can be classified as D-brace (Figure 25a), K-brace (Figure 25b) and V-brace (Figure 25c) according to the shape of the diagonal elements. Eccentrically braced frames belong to the group of dissipative structures and their level of energy absorption is similar to the moment-resisting frame system.

In addition, the eccentrically braced frame system has advantages in terms of drift control. It provides an economic solution in the range of medium and high-rise buildings. The active link is the main energy dissipator in the structural system. It must be designed so that its bending and shear limit strength is reached prior to the attainment of tension and compression limit strengths of other bars.

The length of the active link is responsible for the collapse mechanism which dissipates energy. The short links dissipate energy mainly by inelastic shear deformation in the web (shear link). The long links dissipate energy mainly by inelastic normal strains in the flanges (moment links). A careful design of these links can lead to very satisfactory hysteresis loops with high energy absorption, while maintaining satisfactory rigidity (Figure 26).

Eccentrically braced frames meet the requirements of all three limit states which the seismic design of steel structures considers. In particular they provide excellent strength and rigidity in the elastic range so that non-structural damage and occupant discomfort are avoided. They also have enough ductility to dissipate large amounts of energy in the inelastic range.


• As building height increases, so the dynamic action of wind and seismic loading become more important considerations in design.

• For low-rise buildings, lateral stability can be provided by moment-resisting frames, cross-bracing or shear walls; for high-rise buildings more efficient systems are normally used.

• Four basic categories can be defined - wall, core, frame and tube - and these can be combined to provide more effective bracing systems.

• Appropriate analytical models should be used to determine the performance of lateral bracing systems.

9. REFERENCES

[1] New Structural Systems for Tall Buildings and Their Scale Effects on Cities, Khan, Fazlur R. "Tall Building Plan, Design and Construction", Symp, Proc, Vanderbilt University, Civ Eng Program, Nashville, Tennessee, 1974.

[2] Eurocode Convention of Constructional Steelwork : "Recommendations For Steel Structures in Seismic Zones", ECCS, Publication 54, 1988.

[3] Eurocode 8 : "Structures in Seismic Regions - Design", CEN (in preparation).


1. Steel Designers Manual, Owens G.W.

Blackwell Scientific Publications, Oxford. 1992


OBJECTIVE/SCOPE:

To report the lessons from past failures which may improve the safety of future design and methods of execution.

PREREQUISITES

None

RELATED LECTURES

None.

SUMMARY

In this lecture nine failures are examined. The analysis of these particular cases allows identification of the lack of knowledge or the type of error for the structures concerned. From the analyses, the lessons for future design and execution are highlighted.

Examples of failures due to seismic action are not dealt with, see the lectures of Group 17.

1. INTRODUCTION

In pre-industrial societies, technology and architecture were based largely on a craft approach. The design of objects and buildings changed very slowly over time as gradual improvements were made.

Concepts of progress are not therefore a new idea, but in medieval society builders were restrained to build very carefully, both figuratively and literally, on what had been done before. Failures occurred when they tried to go too far beyond the "state of the art" reached through centuries of slow development.

Cases of failure can be found in the most important and most visible constructions built at that time, the cathedrals. In their desire to have the tallest nave or the widest span in Christendom, the cathedral authorities and builders sometimes strayed beyond the limits of their knowledge and technology. As a result some buildings or parts of buildings collapsed. This was the case with the cathedral of Beauvais. The collapse occurred because the builders had overreached themselves and taken the Gothic structural system beyond its natural limits.

The growth of interest in scientific method and reasoning which started in the seventeenth century led to the industrial revolution. It included development of the ability to predict in advance the forces to which a structure might be subjected when in use. The same process of industrialisation also allowed the production of new materials whose properties were more regular and predictable than those of the natural

materials which they replaced. The combined increase in knowledge and improvement in materials might have been expected to have caused a reduction in the occurrence of structural failures. However, the growth of industrialisation produced a demand for many types of structure for which there was no historical precedent, such as railway stations, covered markets, and exhibition halls.

With the progress of technology came an expectation of novelty from the public and a desire to create it on the part of designers; suddenly it seemed as if almost anything was possible as development accelerated. Each successive structure was for a time the longest, tallest, or had the greatest span. The nineteenth century was a time when designers were faced continuously with trying to solve problems for which there was no precedent.

Without changing demands from society and progress in technology, failures would be caused only by carelessness. By the study of failures, it is possible to learn how to make structures safer as technology develops. This is the subject of the present lecture.

2. ANALYSIS OF SOME STRUCTURAL FAILURES

2.1 General

Failure is by no means the prerogative of ignorance or incompetence. It is more often the consequence of a rare lapse, which team work and vigilance have for once failed to remedy. This lapse may be compounded by ill-luck, by inadequate communication, by safety margins too small to allow for human fallibility, by inexact methods of calculation or construction, etc.

It would be foolish to attempt in a single lecture to make a complete list of reasons for failure and to try to present examples of each. However there are two recurrent themes, most failures occur during erection and one of the most important reasons for structures failing is lack of communication. Poor communication may manifest itself in many different ways. The best guard against it is for all the engineers involved in a job to know each other, to regard each other as friends as well as colleagues involved in a joint enterprise, and most of all to maintain sympathy for one another's views. The difficulty of achieving and maintaining these relationships in a complex contractual situation is discussed in the following section.

Only a very few of the many other reasons for failures recur sufficiently often to warrant specific discussion. Poor detailing may be caused by lack of understanding or by omissions in checking. Numerical error in calculation rarely leads to failure. The inclination to minimise material use, or maximise stresses may also be carried too far, producing only small gain in terms of cost, greater cost in terms of the required accuracy of analysis and/or increased risk of failure. One very clear danger lies in using designs which have proved successful at one scale as a basis for larger structures. The main problem here lies in omissions which were unimportant at the smaller scale becoming significant at the larger scale.

2.2 Contractual Relationship

Usually a job starts with a client who employs an architect to design the structure and control all the other input. The latter will ask a consultant to design the structure. Frequently the consultant produces an outline and member sizes but no joint details. Tender documents will be sent out for the complete structure and each main contractor will ask for subcontract prices for many items of work. Usually the steel frame would be one of these items. In bridgework the architect would not control the work but the main contract/subcontract relationship would still exist.

Once the contracts are let, the steelwork subcontractor will design the connections and begin fabrication. Sometimes even this task is subdivided, with the steelwork subcontractor on the main contract subletting the fabrication work and only doing the erection himself. All of the people mentioned, including "The Engineer"*, who may or may not be the consultant/designer, are bound in a contractual relationship with one another. The contract is very important but is sometimes allowed to disrupt personal relationships between individuals. If ever a breakdown in the friendships between professionals can be seen in a job, then it may be regarded as a clear indication of danger. It is not possible for everyone to carry out their job effectively if there is animosity at any level.

To complicate the problem further, the work is carried out by a labour force which has a corporate identity but which is also a gathering of skilled groups made up of individuals. Safe and economical completion of a job depends on all the members of the team. Mutual respect of skills and interests is needed. If that is maintained, the chances of failure are reduced to negligible proportions. The courage to question the work of others must be matched by a willingness to accept questions and help from others. Similarly the courage to resist pressure for undesirable change is always necessary.

2.3 Structural Failures

2.3.1 Steel box girder bridges

Four failures of steel box girder bridges of somewhat similar design took place during construction in different parts of the world during the years 1969-1971. It is remarkable that no two of these failures were really alike. Two of the bridges were in the cantilever condition when collapse occurred, one of these failed as a result of bottom flange weakness, the other by collapse of a load-bearing diaphragm. One of the other two failed as a result of top flange weakness. The other buckled at the bottom due to temperature differential. All four failures were, however, associated with instability of thin plates in compression.

The main causes of these accidents were:

a. the application of buckling theory with inadequate factors of safety;

b. poor detailing rules and the absence of adequate fabrication tolerances.

The bridges that failed were, in chronological order:

• The Fourth Danube Bridge in Vienna (Austria, 6 November 1969) • The Milford Haven Bridge (United Kingdom, 2 June 1970) • The West Gate Bridge in Melbourne (Australia, 15 October 1970) • The Rhine Bridge in Koblenz (West Germany, 10 November 1971).

A brief description of two of these failures is given below:

Milford Haven Bridge

A local failure during erection of the cantilever on the south side of the bridge led to global collapse. The member concerned was a load bearing diaphragm.

The bridge (Figure 1) was originally designed as a single continuous box girder of welded steel. (It was rebuilt as a cantilever and suspended span in the main span). The spans measured from the south 77m,

77m, 77m, 149m, 213m, 149m and 77m. The span that collapsed was the second 77m span on the south side, the first having been erected with the aid of a temporary support. The collapse occurred when the last section of box for the second span was being moved out along the cantilever. When the collapse occurred this section slid forward down the cantilever killing four men.

It is clear from the reports of the failure that it was initiated by buckling of the support diaphragm at the root of the cantilever being erected (Figure 2). The diaphragm was torn away from the sloping webs near the bottom of the box, allowing buckling of the lower web and bottom flange to take place. As the diaphragm buckled, it shortened, reducing the overall depth of the box girder; the tendency of the bottom flange to buckle was inevitably increased by this reduction of the distance between flanges which increased the force needed in each flange to carry the moment with a reduced lever arm.

The support diaphragm was in fact a transverse plate girder, which carried heavy loads from the webs at its extreme ends, and was supported by the bearings as shown in Figure 2, some distance from its ends. It was therefore subjected to a hogging bending moment and a large vertical shear force. The diaphragm plate near the outer bottom corners was subject to a complex combination of actions. The shear of the transverse girder and diffusion of the point load from the bearings was compounded with the effects of inclination of the webs of the main bridge girder which produced an additional horizontal compression action, and out-of-plane bending effects caused by bearing eccentricity.

The load sustained by the diaphragm just before failure was reported to be nearly 9700kN, which agrees tolerably well with independent calculations of strength made after the accident. The calculated design resistance, using design rules that were drafted subsequently and making allowance for likely values of distortion and residual stresses would be considerably less, possibly as low as 5000kN.

Rhine Bridge, Koblenz

The centre span of the Koblenz bridge over the Rhine collapsed during construction on 10 November 1971, when erection had almost reached the mid-point of the 235m span (Figure 3). The bridge was a single steel box, 16,4m wide at the top plus cantilevers, and 11m wide at the bottom (Figure 4). The box was erected by cantilevering, 85 tons being lifted at a time.

The bottom flange was stiffened longitudinally by T-stiffeners, and the box was stiffened transversely by frames with diagonals made of 300mm diameter steel tubes. All site joints were welded, a comparatively new technique in Germany at the time. As shown in Figure 5, a 460mm gap was provided in the longitudinal T-stiffeners of the lower flange to permit the passage of automatic welding equipment making the transverse butt weld splicing the flange plate. The T-stiffener was then itself spliced by welding in two plates, the plate splicing the web of the T being just 460mm long and butt welded. To avoid a local concentration of residual welding stresses, this plate was not welded to the bottom flange of the box, but was set with its bottom edge 30mm clear of the flange. The plate splicing the table of the T was lapped on top of the ends of the two Ts.

Thus it will be seen that:

• The bottom flange plate, which carried large compressive stresses during construction, was unsupported over a 460mm length at each site splice.

• The main butt weld in the bottom flange plate was at the centre of this 460mm length, possibly introducing a slight lack of straightness.

• The centroid of the splice of the T was almost certainly further from the flange than that of the T itself, thus causing an eccentricity that put the flange plate under a larger compressive stress at this point.

Subsequent investigation showed that the bottom flange plate could carry its stress safely if out-of-straightness was no more than 0,95mm on the 460mm unsupported length. In fact the plate was out-of-straight by as much as 2mm at some points.

On the afternoon of 10 November 1971, preparations to lift the last section of the cantilever from the Koblenz side were complete. Lifting cables were tightened, thus taking part of the weight. A metallic click was heard. The tip of the cantilever settled slightly. A few seconds later the bottom flange splice 50m from the pier buckled and the nose of the cantilever collapsed into the water. The click was undoubtedly the sudden folding up of the flange plate at the splice into the 30mm recess. Much of the stress that should have been carried by the plate was consequently thrown off onto the T stiffeners. They were then taking three times their proper stress, and they buckled too (Figure 6). Thirteen men were killed.

The inquiry concluded that there had been no negligence. The design calculations had all been done correctly according to the methods then normally used in Germany. Rather it was the case that the methods needed revision.

2.3.2 Steel plate girders bridges

King's Bridge in Melbourne

The collapse of the King's bridge in Melbourne is one of the relatively few examples of failure in service. The bridge was opened in 1961 but only 15 months later, on 10 July 1962 (Melbourne's winter), it failed by brittle fracture when a 45 ton vehicle was passing over it. Total collapse was prevented by walls which had been built to enclose the space under the affected span. Investigation showed that many other spans of the bridge were in danger of similar failure.

The foundations were in good order. The superstructure consisted of many spans in which each carriageway was supported by four steel plate girders spanning 30m, topped with a reinforced concrete deck slab. Figure 7 shows a typical girder. The bottom flange of each plate girder consisted of a 400mm x 20mm plate, supplemented in the region of high bending moment by an additional cover plate which was either 300 x 20mm or 360 x 12mm. The cover plate was attached to the flange by a continuous 5mm fillet weld all round.

The steel specified was to comply with BS968: 1941, an earlier version of BS4360 Grade 50 or Fe E 355. BS968 at that time contained no requirement for low temperature notch ductility, but the specification

writer for the bridge did add some special requirements of this type. Despite these additional clauses, those who built and inspected the bridge did not understand that high strength steel needed special care in welding when compared with mild steel (grade 43 or Fe E 275 as it is now called). Difficulties were experienced with welding but an expert was not called in at the time.

Subsequent inspection showed that cracks existed in the main flange plate under 7 of the 8 transverse fillet welds in the span which failed. One crack had extended partly by brittle fracture and partly by fatigue until the tension flange was completely severed, and it had extended half way up the web. All 7 cracks developed into complete flange failures when the failure occurred under a load that was well within the design load. In some instances the entire girder was severed and there was no loss of life. Total collapse was averted by the supporting walls.

Thus the failure of King's Bridge was due to a poor detail which would not be reproduced now, compounded by poor communication which led to a lack of necessary inspection.

Quinnipiac River Bridge

A less dramatic accident occurred in 1973 on the Quinnipiac river bridge near New Haven (USA). A large crack was discovered in a fascia girder of a suspended span.

The structure is non-composite and the girders are 2,8m deep at the crack location. The structure had been in service for approximately 9 years at the time the crack was discovered.

The crack was situated approximately 10m from the west end of the suspended span which is 50m long.

Figure 8 shows the crack that developed in the girder web. The crack propagated to the mid-depth of the girder and had penetrated the bottom flange surface when discovered.

Examination of the fracture surface indicated that crack growth had occurred in a number of stages and modes. These stages are shown schematically in Figure 9.

During fabrication a crude partial penetration weld was placed across the width of the longitudinal stiffener. It is probable that some crack extension from the unfused section occurred during transport, erection and early service. Assuming normal random traffic and approximately 6mm of the 9,5mm thick longitudinal stiffener were unfused, fatigue cracking would require between 2 000 000 cycles and 20 000 000 cycles (depending on the proximity to a free surface) to propagate through the longitudinal stiffener thickness. If the crack had only been fused about 3,8mm on one plate surface so that an edge crack resulted, only about 1 000 000 cycles of random traffic would be needed to propagate the crack through the longitudinal stiffener. Fatigue crack growth (Stage II) would develop mainly after the stiffener was cracked in two. Electron microscope studies of the fracture surface confirmed the presence of fatigue crack growth striations during stage 2.

Stage 3 was the brittle fracture of the web during a time of low temperature. It was initiated in a zone of high residual tensile stresses. Once the crack became unstable, it propagated through the zone of lower stresses in the web and was eventually arrested in the flange. Further fatigue crack growth (Stage 4) developed thereafter and continued until the crack was discovered and repaired.

In this case the failure was due to a weld containing an internal defect (lack of fusion) which initiated a fatigue crack. Total failure was avoided by the detection of the crack during regular inspection.

Bridge on the Sainte Marguerite River in Sept-Iles (Quebec)

As in the Milford Haven bridge a local failure led to global collapse. The bridge on the Sainte Marguerite River consisted of five steel plate girders made composite with the deck. As shown in Figures 10 and 11, each girder had four supports, two on the abutments and two on cross-beams joining the top of the inclined legs at an angle of 45° strut.

The cross-beams were supported by a group of three braced struts at the same inclination. The bridge deck consisted of a concrete slab (220mm) with an asphalt layer of 65mm. The composite behaviour was provided by stud connectors welded on the steel girders.

The bridge failed during the asphalt surfacing. The failure was initiated by a local buckling of the webs of the struts on the Sept-Iles side (Figure 12). The support provided by these struts vanished and, as a result, the span increased from 54,0m to 95,8m. The bending moment in the main girders was multiplied by a factor of 5. The composite girders and deck were not able to resist and failed.

The main reason for the failure was found in the assembly between cross-beam and struts. Without stiffeners the webs of the struts (WWF 900 x 293) were too slender and not able to resist the axial loads transmitted by the cross-beam. The width to thickness ratio of the webs had a value of 76,7 while the

limiting value is about 34 for the steel considered (following Eurocode 3: b/t ≤ 42 ε and ε = ). In this condition, the maximum axial load which could be carried by the strut was about 3300 kN, a value later confirmed by test. The strut load at the moment of failure had a value of 3500kN while the calculated service load was 5780kN.

To carry the service loads with a reasonable safety factor it was necessary to place stiffeners on the web of each strut in order to obtain full collaboration of the web (Figure 13).

This failure can be attributed to an insufficient knowledge of the behaviour in compression of struts with slender webs.

2.3.3 Shell structures

The Seneffe Water Tower (Belgium)

A type of steel water tower which is rather popular in Belgium and abroad, is shown in Figure 14a. The main shell, where the water is stored, is theoretically axisymmetrical about the vertical axis and is often compared to a golf ball with its Tee support. Such a water tower with a capacity of 1500 cubic metres was built in 1972 near the industrial park of Seneffe; the main dimensions are shown in Figures 14a. and b. Two conical shells made of 8 and 15mm thickness steel formed the main part of the water tower.

The initial design was made using membrane theory. Assessment of local bending stresses at the shell intersection was derived from an axisymmetrical analysis using finite elements. Both computation methods were of the first order but they did not take account of any instability phenomena. Indeed, due to internal pressure, the hoop stresses in the part AB of the water tower (which were found later to be critical) are tensile; in spite of the fact that the meridional stresses in the upper part are compressive, no analysis of a possible buckling of the conical shells was undertaken.

During the first filling test the water tower collapsed when the water level corresponded to a volume of 1130 cubic metres, i.e. when the water had risen to 1,74m below the overflow level (Figure 14). Failure

occurred by buckling on the thin cone near the junction of the two cones. The collapsed structure is shown in Figure 15.

After the accident the available literature was carefully scrutinized and the following conclusions emerged:

a. Imperfections, which may be geometrical or structural. The welding procedure, used quite generally to assemble the various components of a branched shell such as that shown in Figure 14, produced both local geometric imperfections and high residual stresses. The residual stresses are never reduced by an annealing treatment except in steel nuclear vessels.

b. Discontinuity stresses, which have high local peaks at the intersections of branched shells.

At that time, even using the most advanced information on stability of isolated shells, it was only possible to have an idea of the collapse resistance of a very perfect shell under idealised boundary conditions. The available literature disregarded entirely the effect of imperfections and discontinuity stresses.

The failure of the Seneffe water tower was the starting point for important experimental research in the fields of liquid-filled conical shells and of nonlinear computer analysis taking into account geometrical imperfections.

The last edition (1988) of the ECCS Recommendations on the Buckling of Steel Shells gives much information for a wide range of cylindrical, conical and spherical shells. Current design recommendations relating to the buckling of manufactured shell structures now take account of realistic levels of geometric imperfections and residual stresses.

Wind effects on a steel chimney

After a five-year service period a 25,81m high steel chimney in a group of four chimneys partially failed during a storm with windspeeds of between 120 and 150km/h.

The four chimneys consisted of conical and cylindrical shells of 800mm diameter assembled by bolted flanges or by welding (Figure 16). In the bolted connection of an external flange located at a height above ground of 13,575m, 13 of the 24 bolts broke. The deformation of the flange led to a perceptible slope of the upper part of the chimney, which did not fall.

At the moment of the accident the wind blew along the line of the four chimneys, from West to East. The fourth chimney (down wind) was damaged and the failure of the bolts affected the South part of the flange. This position corresponded to bending of the chimney perpendicular to the wind direction which is a characteristic of vortex shedding.

Where cylinders are in line, the effects of vortex shedding are greater on and after the second cylinder than on an isolated cylinder when the distance from axis to axis is less than 10 diameters. The effects can be doubled.

The forces on the bolts due to the bending perpendicular to the wind direction were unfortunately underestimated and the flanges were not stiff enough. The failure of the bolts was due to fatigue in bending and was initiated in an overloaded bolt. The overloading was due to inadequate tightening of adjoining bolts and to deformation of the flange.

The main reason for the failure was because the amplification of vortex shedding in the case of chimneys in line was not taken into account. In Eurocode 1: Basis of Design and Actions on Structures, special attention will be given to the additional dynamic effects of wind on structures.

2.3.4 Buildings

Zoology Block, Aberdeen University

The Zoology Block was a rectangular six-storey building with a steel frame and a plan area of 13 metres x 65 metres. The steel columns were placed along both sides of the building at 2,82 metre centres and carried steel beams 686mm deep which spanned the full 13 metres. Figure 17 shows the simple angle brackets that were fixed to the external face of the columns and supported horizontal 152mm x 152mm steel universal column sections; these were to carry the proposed cladding of precast concrete panels. The floors were precast concrete planks bearing directly onto the 13 metre span main beams.

After erecting the steelwork, the contractor decided to install the floors to facilitate later work. Unfortunately, with the additional weight of the precast concrete planks, the columns of the building were on the point of buckling in the long direction of the building and a light wind provided the necessary disturbing force. The whole building collapsed in the longitudinal direction with the floors stacked one above the other. Four men were found dead in the wreckage and one died later. Several others were injured.

The wind was not unusually strong on the day of the collapse, but it was enough to put the building out of plumb. Once the structure moved out of the vertical, the mass of the concrete floors created a large overturning moment. The only stiffness in the plane of the collapse was in the cleats which attached the cladding rails to the frames formed by the main columns and beams. The cladding rails passed outside the plane of the columns and the fixing cleats were virtually pin joints and allowed the rails to rotate relative to the columns.

The main reason for the collapse was the lack of sway restraint, which would have been provided once the cladding panels were in place to stiffen the structure. It could be concluded therefore that the mistake lay in the process of erection; the cladding should have gone in before the floor panels. However the contractor had no reason to assume that the steel frame could not support all the possible loads applied to the building. He should have been told if this was the case. This would mean that the mistake was one of communication. If the connections between the cladding supports and the columns had been designed to be somewhat stiffer, the collapse would have been unlikely to have happened. Tests and calculations made subsequently showed that collapse of the building in the long direction was much more likely than collapse in the short direction, although it is the latter, with the longer face exposed to the wind load, that would generally be thought more likely. Under the steelwork design code, Eurocode 3, the contractor should be informed if the steel frame is not stable in its own right before the cladding is put on it, so that he can then plan the erection sequence accordingly.

During the investigation into the collapse, it was discovered that the original cleat detail for the cladding rails, which would have provided some stiffness in the plane of the collapse, had been revised when erecting the frame because the cleats were "awkward" to construct on site. In the end both the designers of the steel frame and the contractor who erected it were found liable for the collapse because of their "most unfortunate and quite unintentional misunderstanding" due to lack of communication.

Hyatt Regency Hotel, Kansas City

On the 7 July 1981, a dance was being held in the lobby of the Hyatt Regency Hotel, Kansas City. As spectators gathered on suspended high level walkways, the supports gave way and two levels of bridge fell to the crowded dance floor. One hundred and eleven people died and nearly two hundred were seriously injured. Failure occurred at a simple but critical detail.

The walkways crossed the lobby at second and fourth floor levels and were supported above one another by hanger rods from the fifth floor (Figure 18). Floor to floor height was 5m and the walkways hung from three sets of hangers at 9m centres. In the original design single 15m long rods supported the two walkways (Figure 19(a)). At each level a cross-beam made from two channels welded toe to toe rested on a nut and washer on the rod. This detail would not have failed under the loading imposed even though its strength was only one quarter of that required by local design codes.

In the furore which followed the collapse it became obvious that the design had been changed to reduce the cost of the connection. The second floor walkway was actually suspended from the fourth floor one (Figure 19(b)). As a result the connection between the fourth floor cross-beam and the hanger supported double the load originally intended, or rather failed to do so. The alteration seems to have been recommended by an engineer, not party to the original design, who specialised in reducing costs.

Unfortunately he failed to understand the importance of the details changed: nor was the effect of the changes spotted by any of the other parties involved.

Once again, a gross underdesign of a detail would not have caused failure had not another factor resulted in a significantly increased load. Here, as in most failures, lack of communication was the most important reason why failure was not prevented.


• Society rightly demands a high standard of safety from civil engineering structures. When a structure fails, it may claim many lives, and its reinstatement may require considerable resources.

• Structures rarely fail from a single cause; there are usually several contributory factors to failure. • Structures are frequently at greatest risk during construction. • The structures described in this lecture failed from one or more of the following causes:

⋅ poor communication.

⋅ design error or lack of understanding of structural behaviour.

⋅ a material-related problem causing failure in a structure even though its behaviour is reasonably well understood by the designer.

⋅ errors in detailing or poor detailing rules caused by lack of understanding or checking.

⋅ inadequate temporary works, lack of thought about a temporary condition or about the process of erection.

• Failure is by no means the prerogative of ignorance or incompetence. Even in routine work according to recognised codes, failure is more often the consequence of a rare lapse which team work and vigilance have for once failed to remedy. This lapse may be compounded by ill-luck, by inadequate consideration of the fundamental behaviour of the proposed structure, by safety margins too small to allow for human fallibility, and by inexact methods of calculation or construction.

• The study of some accidents stimulates research such as, for example, into the buckling behaviour of plate and shell structures. Some case histories indicate the need to review the bases of codes or design methods.

• Success or failure is ultimately the work not of codes but people; success depends primarily on the engineer and his team.


1. Smith, D. W., Bridge Failure, Proc. Instn. Civ. Engrs., Part 1, 1976, 60, August, pp 367-382. 2. Roik, K., Betrachtungen über die Bruchursachen der neuen Wiener Donaubrücke, Tiefbau, Vol.

12, p 1152, 1970.

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Steel Construction Introduction to Design

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