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guidelines for
design of dams
for earthquake
- lu#_,
AUSTRALIAN NATIONA L
COM MITTEE ON
LAR GE DAMS
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GUIDELINES FOR
D E S IG N O F D A M S
FOR EARTHQUAKE
AUG UST 1998
AUSTRA LIAN NATIONA L
COM MITTEE ON
LARGE DAM S
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A U S T R A L IA N N A T IO N A L C O M M IT T E E O N L A R G E D A M S
G U ID E L IN E S FO R D E S IG N O F
D A M S FO R E A R T H Q U A K E
AUGUST 1998
IM P O R T A N T D IS C L A IM E R
"ANCOLD and its Members, and the Convenor, Members and Assistants of the
Working Group which developed these Guidelines do not accept responsibility
for the consequences of any action taken or omitted to be taken by any person,
whether a purchaser of this publication or not, as a consequence of anything
contained in or omitted from this publication. No persons should act on the basis
of anything contained in this publication without taking appropriate professional
advice in relation to the particular circumstances".
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T A B LE O F C O N T E N T S
Page No.
F O RE W O RD
ANCO LD WORKING GRO UP MEMBER SHIP L IST
NTRODUCTON1
EARTHQUAKES AND THEIR CHARACTERISTICS2
2.1Earthquake Mechansms and Termnoogy2
22EarthquakeGoundMoon2
23SuaceRupue3
24MagntudeandInensty3
2.5Changes to Seismic Waves Near the Ground Surface4
2.6Attenuation and Amplification of Ground Motion4
27Reservor InducedSesmcty5
EARTHQUAKE HAZARD IN AUSTRALIA 6
31Gnea6
32Mechansmo Earhquakes8
33EahquakeDphs9
34Evauaono Sesmc Hazad10
35Aenuaon11
3.6MaximumCredbe Earthquake Magntude11
3.7Estimates of Ground Motion and Response Spectra at a Site12
38EarthquakeHazardMaps12
SELECTION OF DESIGN EARTHQUAKE 16
41Denons16
42Seection of the Design Earthquake20
4.3 Selection of the Operating Basis Earthquake (OBE) 32
44Concurrent Load Combinations32
45Earthquakes Induced by the Reservor33
46Response Spectra and Acceerograms *33
D E S I GN OF E M BA N K M E N T D A M S A N D A N A LY S IS OF
LQUEFACTON33
5.1 Effect of Earthquake on Embankment Dams3 3
5.2 General ("Defensive") Design Principles for Embankment Dams 34
5.3Liquefaction of Dam Embankments and Foundations36
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6. SEISMIC STABILITY ANALYSIS OF EMBANKMENTS 57
61Peame57
62Pseudo-Sac Anayss57
6.3 Simplified Methods of Deformation Analysis59
6.4Post Liquefaction Stability and Deformation Analysis63
65Numerca Mehods65
66PoposedGudenes67
7.ANALYSIS AND DESIGN OF CONCRETE DAMS69
7.1 Past Performance of Concrete Dams in Earthquakes69
72DeensveDesgnMeasues70
73AnayssMhods71
7.4Design Earthquake and Hydrodynamic Loads82
75DsgnCea83
76Dynamc Maea Popetes86
8APPURTENANT STRUCTURES87
81noducon87
82PerormanceRequremens87
83nakeTows89
RE F E RE NC E S
APPEND IX A
TERMS OF REFERENCE
APPENDIX B
TYPICAL EASTERN AUSTRALIAN PEAK
G RO UND A C C E L E RA T IO N V S A E P
RE S PO NS E S PE C TR UM FO R 1 in 1000 A E P
MODIFIED MERCALLI SCALE
APPENDIX C
EXTRACTS FROM CANADIAN DAM SAFETY
GUIDEL INES
APPENDIX D
ADDITIONAL INFORMATION ON ACCEPTABLE
RISKS
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F O R E W O R D
Even in the matter of earthquakes, Australia can be considered the "Lucky Country" in not being on the
edge of major tectonic plates. Neighbouring countries like New Zealand and Indonesia are renowned for
their volcanoes and frequent earthquakes. Australia is relatively earthquake free by comparison and
earthquakes were seldom considered in early dam designs.
Certainly there were some zones of known activity such as the Adelaide Hills and the Western
Australian wheat belt, and whilst major damage had occurred, it was not on the same scale as in other
countries.
In 1979, the Standards Association of Australia produced the "Earthquake Code" AS2121. It showed
zones of seismic activity and recommended methods of determining loads on building structures. The
development of this code was based largely on statistics of historic earthquakes, for which there were
relatively short term records.
However, several major earthquakes subsequently occurred in areas indicated by the code as having
negligible earthquake risk, the most notable being the 1989 earthquake at Newcastle (Magnitude 5.6) in
which 12 people died and the Tennant Creek Earthquake in 1988 (Magnitude 6.8). This led to the
introduction of a new earthquake code (AS 11 70.4-1993)which included data from m ore widespread and
reliable seismographs and furthermore considered the all important geological situations.
In parallel with these developments, analytical methods used by dam engineers were improving beyond
the simplistic application of a horizontal force equating to seismic acceleration. Improvements were
based on the observed fact that earth dams subjected to earthquakes had slumped vertically rather than
fail by slipping of a face as indicated by the simplistic analyses.
Methods of analysing slumping were developed, and further supplemented by sophisticated finite
element analyses which, by utilising modem computer power, give an ability to undertake rigorous
analyses of dams where necessary.
This ANCOLD Guideline brings together improved appraisals of the earthquake loadings that a dam
may suffer and then describes appropriate methods for analysis and evaluation. Whilst specific to the
Australian considerations, the majority of this guideline could be applied to dam structures throughout
the world. The IC O LD Bulletins No. 46(198 3), No. 52(198 6), No. 62(198 8) and No. 72(1989 ) are
parallel documents in this regard, although not including recent advances.
This guideline is a major contribution to dam engineering and the voluntary work by the ANCOLD
subcommittee has unselfishly provided their experience to the dam building community and indeed the
wider community. Our appreciation goes to Prof Fell and his team for producing this valuable guideline.
This guideline is not a design code, and dam designers must continue to apply their own considerations,
judgements and professional skills when designing dams to resist earthquakes. As time goes on there
will no doubt be improved data and design tools to help the designer and it is intended that this guideline
will be updated as circumstances dictate. ANCOLD welcomes contributions to discussion on this
guideline which will assist with future revisions.
/
JOHN PHILLIPS
Chairman, ANCOLD
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MEMBERSHIP OF THE ANCOLD WORKING GROUP FOR
G U ID E L IN E S F O R T H E D E S IG N O F D A M S F O R E A R T H Q U A K E
4
Robin Fell
School of Civil Engineering, University of New South Wales
Gamini Adikari
Snowy Mountains Engineering Corporation, Victoria
John Bosler
Snowy Mountains Engineering Corporation, Cooma, New South Wales
Brian Cooper
Dam s and C ivil Section, Public W orks and Services Department, NS W
Peter Foster
Works Consultancy Services, Power Engineering, New Zealand
Gary Gibson
Seismology Research Centre, RMIT, Melbourne
Sergio Giudici
Hydro-Electric Commission, Hobart
Nasser Khalili
School of Civil Engineering, University of New South Wales
Ian Landon-Jones
Dams S afety Group, Sydney Water, New S outh Wales
Kevin McCue
Australian Seismological Centre, Canberra
Len McDonald
Dams and Civil Section, Public Works and Services Department, NSW
Brian Shannon
Water Resources, Department of Primary Industries, Queensland
David Stapledon
Geotechnical Consultant, Adelaide, South Australia
John Waters
Geo-Eng Pty Ltd, Perth, Western Australia
Ron Wyburn
Halcrow Water Power, Victoria
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1. INTRODUCTION
Public awareness of the potential for damage
and loss of life in Australia from earthquakes
was highlighted by the Newcastle earthquake in
December 1989. This was a Magnitude 5.6
(M5.6) event, and because of its proximity to
Newcastle, and local ground conditions, caused
approximately $1 billion damage.
Dam engineers in Australia have been
conscious of earthquakes for many years, but it
was the earthquakes at Tennant Creek in 1988,
which were M6.3, M6.4, M6.7, with a total fault
scarp length of 32km which raised the question
most acutely as to whether dams in Australia
could be subject to large earthquakes, and if so,
could they withstand them without resultant
loss of the facilities and lives, property, and
environmental values downstream. Other large
earthquakes in the M6 to M7 range had
occurred in Australia, the most notable being in
Meckering in 1968 (M6.9), but the Tennant
Creek event was critical because it occurred in
an area which had previously been regarded as
virtually free of earthquakes.
Recent assessments of earthquake ground
motions for some large Australian dams have
been based on the assumption that the
maximum credible earthquake is M7.5, which is
large by any standards. The seismologists
involved in these studies indicate that on the
available evidence, such earthquakes, ie. M7.5,
could occur anywhere in Australia.
In general, it is not possible to identify active
faults which might cause such earthquakes. For
example, there had been no movement on the
Tennant Creek fault for more than 200,000
years (Crone and Machette, 1992), so the
question arises, can it occur at, or close to any
damsite? Peak ground accelerations close to a
M 7.5 earthquake can be very high.
The past performance of dams in earthquake
h&s been very good, with few dams suffering
major damage. Where this has occurred, it has
been due to liquefaction in the dam or the
foundation. Very few of these dams have
breached and released a flood wave. However,
several might have breached, if the reservoir
level had been higher at the time of the
earthquake. Seed (1979), USCOLD (1992),
ICOLD (1986), NSWDSC (1993) and Hinks
and Gosschalk(1993) give some details.
The approach taken by seismologists in
Australia is to use statistical analysis to predict
the frequencies of recurrence of ground
motions. This typically results in 1 in 1000
AEP peak ground accelerations of 0.15g, 1 in
10,000 AEP s0.35g, and 1 in 100,000 AEP
~0.5g. These are large loadings and it is likely
that assessment of many of the existing dam s in
Australia for such loads could indicate some
deficiencies to either the dam or appurtenant
structures. New dams would also need (less)
expensive additional design features to cope
with earthquake.
In recognition of the need to provide some
guidance to dam engineers and owners in
Australia, ANCOLD established a Working
Group to prepare Guidelines for the Design of
Dams for Earthquake. The Working Group was
established in September 1993, and took over
from an earlier ANCOLD Working Group
preparing Guidelines on Seismic Analysis and
Design of Embankment Dams. The Terms of
Reference for the Working Group are in
Appendix A.
These guidelines are to cover all types of dams,
including tailings dams, and apply to existing
and new dams. They cover the selection of the
design earthquake, analysis and design of
embankment and concrete dams, and
appurtenant structures.
The guidelines are not meant to be used as a
design code, and of necessity, do not include
complete details of all the analysis and design
methods which are recommended. The area is
rapidly evolving, and those involved in the
analysis and design of dams for earthquake
should refer to the references given, and to
more recent publications so as to be fully
informed. In some situations it will be
necessary to seek specialist advice.
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2. EARTHQUAKES AND
T H EIR CH A R A CT ER IS T ICS
2.1 Earthquake Mechanisms and
Terminology
An earthquake is the motion that is produced
when stress within the earth builds up over a
long period of time until it eventually exceeds
the strength of the rock, which then fails and a
break along a fault is produced. It may take
tens, hundreds or thousands of years for the
stress to build up in a particular area, and it is
then released in a few seconds. Part of the
energy is transmitted away as seismic waves
and part of it as heat.
The fault displacement in a particular
earthquake may vary from centimetres up to a
few metres in a great earthquake. Once
ruptured, the fault is a weakness which is more
likely to fail in future earthquakes, so a large
total displacement may build up from many
earthquakes over a long period of time. This
may eventually measure kilometres for thrust
faults produced by compression, or hundreds of
kilometres for horizontal strike-slip faults such
as the San Andreas.
The point on the fault surface where a
displacement commences is called the
hypocentre or focus, and the earthquake
epicentre is the point on the ground surface
vertically above the hypocentre. The
displacement usually propagates along the fault
in one direction from the hypocentre, but
sometimes it propagates in both directions.
Energy release is near but not exactly at the
hypocentre.
The hypocentral distance from an earthquake to
a point is the three dimensional slant distance
from the hypocentre to the point, while the
epicentral distance is the horizontal distance
from the epicentre to the point.
2.2 Earthquake Ground Motion
Earthquake ground vibration is recorded by a
seismograph or a seismogram. Most modem
seismographs record three components ol
motion: east-west, north-south and vertical.
The rupture time for small earthquakes is a
fraction of a second, for earthquakes of
magnitude 5.0 it is about a second, and for large
earthquakes may be up to tens of seconds.
However the radiated seismic waves travel al
different velocities, and are reflected and
refracted over many travel paths, so the total
duration of vibrations at a site persists longei
than the rupture time, and shows an exponentia
decay.
Several types of seismic wave are radiated from
an earthquake. Body waves travel in three
dimensions through the earth, while surface
waves travel over the two dimensional surface
like ripples on a pond. There are two types oi
body wave (P and S waves), and two types oi
surface wave (Rayleigh and Love waves).
Primary or P waves are ordinary sound waves
travelling through the earth. They are
compressional waves with particle motion
parallel to the direction of propagation.
Secondary or S waves are shear waves, with
particle motion at right angles to the direction of
propagation. The amplitude of S waves from an
earthquake is usually larger than that of the P
waves.
P waves travel through rock faster than S
waves, so they always arrive at a seismograph
before the S wave.
The frequency content of earthquake ground
motion covers a wide range of frequencies up to
a few tens of hertz (cycles per second). Most
engineering studies consider motion between
about 0.2 and 25 Hz.
The amplitude, duration and frequency content
of earthquake ground motion at a site depend on
many factors, including the magnitude of the
earthquake, the distance from the earthquake to
the site, and local site conditions.
The larger the earthquake magnitude, the
greater the amplitude (by definition a factor of
ten for each magnitude unit), the longer the
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duration of motion, and the greater the
proportion of seismic energy at lower
frequencies. A small earthquake has low
amplitude (unless it is very close), short
duration, and has only high frequencies.
The smaller the distance from an earthquake to
the site, the higher the amplitude. The duration
is not strongly affected by distance. High
frequencies are attenuated by absorption within
the ground more quickly than low frequencies,
so at greater distances the proportion of seismic
vibration energy at high frequencies will
decrease.
2.3 Surface Rupture
Surface rupture is a relatively rare phenomenon
which occurs when a fault break reaches the
ground surface. It may produce a vertical or
horizontal offset (or both) with a displacement
of millimetres to a few metres, and a length
from metres to tens of kilometres.
Because rock near the surface is relatively
weak, few earthquake hypocentres occur in the
top one or two kilometres. It is common for
surface sedimentary rocks to be folded in
response to faulting at depth, giving a
monocline and scarp at the surface, but without
a surface fault.
Most earthquakes, especially most larger
earthquakes, occur on existing faults. This is
because faults are weaker than surrounding
unbroken rock, and are much more likely to fail
again when stress rebuilds.
A site will have surface rupture potential if an
existing fault is found which has been active in
the recent geological past (perhaps the past few
million years). This will be quite rare, and
possibly be difficult to establish. It will usually
be easier to show that a site with simple surface
geology has no faulting history, than to show
that a site with complex geology has suffered
recent faulting.
2.4 Magnitude and Intensity
Earthquakes vary enormously in size. In 1935
Richter defined a magnitude scale to indicate
the size of an earthquake. For the Richter local
magnitude scale, ML, the logarithm of the peak
ground displacement is taken and an empirical
correction depending on the distance from
earthquake to seismograph is subtracted. The
resulting values are averaged for all the
seismographs that have recorded the
earthquake.
Other magnitude scales have been defined,
including moment magnitude, and while not
exactly the same as the R ichter local magnitudes,
they give similar values that can range from 0.0
to over 9.0. For each unit of magnitude there is
a tenfold increase in ground displacement, and a
thirtyfold increase in seismic energy release.
Another measure of earthquake size is the fault
area, or the area of the fault surface which is
ruptured. The fault area ruptured in an
earthquake depends on the ma gnitude and stress
drop in the earthquake. For a given magnitude,
a higher stress drop will give a smaller rupture
area. Typically, a magnitude 4.0 earthquake
ruptures a fault area of about 1 square
kilometre, magnitude 5.0 about 10 square
kilometres, and magnitude 6.0 about 100 square
kilometres (perhaps 10 by 10 kilometres).
Earthquake Intensity is a measure of the effect
of the seismic waves at the surface, and is
normally given on the Modified Mercalli
Intensity scale, a copy of which is attached in
Appendix B. This is an arbitrary scale defined
by the effects observed (whether sleeping
people were woken, trees shaken, etc) and on
the amount of damage caused. Normally the
maximum intensity occurs near the epicentre of
the earthquake, and intensity then decreases
with distance. However, this may be affected
by the orientation of the earthquake rupture, or
by local ground conditions such as topography
or surface sediments.
The earthquake recurrence or seismicity
(seismic activity) of an area must take the range
of earthquake sizes into account. There are
many more small earthquakes than large. In
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most places around the earth there are about ten
times as many earthquakes exceeding
magnitude 3.0 than there are exceeding
magnitude 4.0, and ten times as many again
exceeding magnitude 2.0. In seismicity studies,
the logarithm of this factor is called the b value,
so a value of 1.0 is typical. The b value may be
1.3 or higher if there are many small
earthquakes, or 0.7 or lower if there are few
small earthquakes.
2.5 Changes to Seismic Waves Near
the Ground Surface
The energy in seismic waves depends upon
their amplitude and the physical properties of
the material through which they are passing.
When waves pass from high stiffness material
(eg. rock at depth) into lower stiffness material
(eg. near-surface rock, or sediments) they are
reflected towards the vertical and their
amplitude increases. Their amplitude also
increases as they approach the earth's (free)
surface, at which they are reflected. The nature
and extent of free surface amplification varies
with topography, even in fresh, strong rock.
Changes in soil thickness above an irregular
bedrock surface can give complex surface
amplification that varies with earthquake wave
duration.
Resonance in the surface sediments causes
amplification at particular frequencies,
especially at the natural frequency of the
sediments. This depends on the thickness and
elastic properties of the sediments. Earthquake
motion recorded on hard rock includes all
frequencies up to a value that depends on
magnitude, while that recorded on soft
sediments is usually dominated by the resonant
frequency.
In surface sediments, high frequency vibrations
are attenuated much more with distance than
low frequencies. If sediments are very thick,
much of the high frequency motion will be lost
and peak surface accelerations will be low, even
if resonance has amplified motion at the low
resonant frequency.
2.6 Attenuation and Amplification of
Ground Motion
Earthquake ground motion attenuates with
increasing distance from the source due to
radiation and hysteretic damping. High
frequency motion is attenuated more quickly
with distance than lower frequency motion.
For estimates of peak ground acceleration,
attenuation is allowed for by using an
attenuation function of the form
a =b,eb2Mr*3
where a =acceeration
R=foca dstance
M=Magntude
b jb are constants, which
vary considerably over the
world.
Some earthquake hazard studies use the Esteva
and Rosenblueth (1969) attenuation functions,
which give peak ground velocity (mm/s), peak
ground acceleration (mm/s2) and Modified
Mercalli Intensity (IMM) at an epicentral distance
x kilometres from an earthquake at depth z
kilometres with local magnitude M. The
equations are:
R
=
Vx2 +z2 +400
Vpeak
= 160 e10 M R"17
Speak
=
20000 e0 8 mR20
Im m
=
loge(2980e15MR-23)
Because of the 400 term in the expression for R,
corresponding to a minimum R of 20
kilometres, these equations give low values of
ground motion at distances closer than a few
kilometres.
These relations were determined using
Califomian data, and should only be used with
magnitudes determined using a compatible
function. If the magnitudes computed for
seismographs at different distances vary, then
the attenuation function is invalid for the area .
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In selecting attenuation relationships care is
needed, and attention paid to the mechanism of
the source earthquake, eg. whether shallow
intraplate, or deep crustal boundary
earthquakes.
Weak surface materials absorb seismic energy
rather than transmit it unchanged, thus tending
to reduce amplitudes at the surface. The
amount of attenuation depends on the properties
of the materials, and especially on their
thickness.
Near-surface layers will vibrate preferentially at
their own natural frequencies, depending on
their thickness and elastic properties. The
earthquake motion at the natural frequencies of
the near-surface layers is amplified, while
motion at other frequencies may be little
affected or even attenuated. The amplification
effect can be especially pronounced for deep
soft sediments such as those underlying M exico
City, but in deep, stiff sediments subject to high
frequency earthquake, attenuation may result.
Dams (like all other structures) have natural
frequencies of their own depending on their
mass and stiffness, usually in the range from
about 0.5 hertz to about 5 hertz for embankment
dams and 2 hertz to 20 hertz for concrete
gravity dams.
2.7 Reservoir Induced Seismicity
Reservoirs may induce seismicity by two
mechanisms. Either the weight of the water
may change the stress field under the reservoir,
or the increased ground water pore pressure
may decrease the stress required to cause an
earthquake. In either case, reservoir induced
seismicity (RIS) will only occur if relatively
high stresses already exist in the area. If the
stress has been relieved by a recent large
earthquake, say in the last few hundred years
for low seismicity areas like Australia, then RIS
is unlikely to occur.
/
RIS events initially usually occur at shallow
depth under or immediately alongside a
reservoir. As years pass after first filling, and
groundwater pore pressure increases permeate
to greater depths and distances, the events may
occur further from the reservoir. This occurs at
a rate of something like one kilometre per year.
RIS is experienced under new reservoirs,
usually starting within a few months or years of
commencement of filling, and usually not
lasting for more than about twenty years. Once
the stress field and the pore pressure fields
under a reservoir have stabilised, then the
probability of future earthquakes reverts to a
value similar to that which would have existed
if the reservoir had not been built. Most of the
earthquake energy does not come from the
reservoir, but from normal tectonic processes.
The reservoir simply acts as a trigger.
In areas with horizontal tectonic compression
and reverse faulting, like Australia, filling a
reservoir should increase the vertical minimum
principal stress and reduce the chance of an
earthquake under the reservoir. This has been
called reservoir induced a seismicity. However,
in some cases earthquakes could then be
induced by later releasing water from the
reservoir. Alternatively the change in stress
during filling could induce earthquakes beside
the reservoir rather than under it, although this
stress change is less pronounced.
It has been suggested that filling a reservoir will
cause compression under it, increasing the pore
pressure of the existing groundwater, and so
tend to induce earthquakes even in areas of
horizontal compression. Stress change induced
seismicity, either direct or through this indirect
mechanism, should occur soon after filling. It
may then cause seasonal variations in
seismicity, sometimes lagging a few weeks or
months behind water level.
Pore pressure induced seismicity is normally
delayed, and may occur years after filling. Pore
pressure increases always tend to induce events.
If there is a major fault near the reservoir, RIS
can produce earthquakes exceeding magnitude
6.0 (Xinfengjiang, China, 1962, M6.1; Koyna,
India, 1967, M6.3). Such events will only occur
if the fault is already under high stress. A
number of Australian reservoirs have triggered
earthquakes exceeding magnitude 5.0
(Eucumbene, 1959, M5.0; Warragamba, 1973,
M5.0; Thomson, 1996, M5.2).
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MPP*8-
It is more common for a reservoir to trigger a
large number of small shallow earthquakes,
especially if the underlying rock consists of
jointed crystalline rock like granite (Talbingo,
1973 to 1975; Thomson, 1986 to 1995). These
events possibly occur on joints rather than
established faults, so are limited in size, and
only give magnitudes up to 3 or 4. There is no
hazard from such low magnitude reservoir
induced earthquakes, even if they occur
regularly. Their shallow depth means that they
may often be felt or heard.
RIS has been observed for over one hundred
reservoirs throughout the world, and small
shallow induced events have probably occurred
under many others. A relatively high
proportion of reservoirs with RIS seismograph
networks do record such activity. A high
proportion of RIS examples occur in intraplate
areas, with above average rates in China,
Australia, Africa and India.
It is not easy to predict whether a reservoir will
experience RIS because the stress and strength
at earthquake depths are not easily measured.
For the same reason, prediction of normal
tectonic earthquakes has been unsuccessful in
most parts of the world.
It seems that RIS with many small events is
more likely to occur in intraplate areas with
near surface crystalline rocks like granite, rather
than sedimentary rocks. A larger magnitude
RIS event can only occur if there is an existing
fault of sufficient dimension that is late in its
earthquake cycle (the stress is already
approaching the strength of the fault).
3. EARTHQUAKE HAZARD
IN AUSTRALIA
The understanding of the hazard imposed by
earthquakes in Australia is critical to selection
and application of design earthquakes for dams.
Hence, a relatively detailed discussion on the
topic follows. This is largely taken from Gibson
(1994).
3.1 General
The Australian continent is within a tector
plate shared with Southern India, so all of
earthquakes are intraplate. The pla
boundaries to the north and east are among t
most active on the earth. Possibly as a result
this, Australia is one of the most actr
intraplate areas on the earth. Despite this, tf
hazard is quite low when compared with acth
interplate areas.
Most people in Australia can expect to feel
earthquake about every five or ten year
although many of these may not be recognisf
as an earthquake. Most Australian earthquake
that are reported are heard with a noise like
distant quarry blast or thunder, with possibly
slight vibration being felt.
Only a proportion of earthquakes that are fel
perhaps about one in twenty, will cause som
damage in their epicentral area. If they occur i
an inhabited area, most earthquakes larger tha
about magnitude 4.0 will cause some damage.
By contrast, in an active interplate area lik
New Britain or Bougainville in Papua Nev
Guinea, earthquakes are felt very often, c
average every week or two. These are normal
felt rather than heard, with any soun ds being thf
reaction of a building to the vibration rathe
than the earthqua ke itself.
PNC
havf
A very small proportion of these
earthquakes, perhaps about 1 in 500,
caused any damage in their epicentral area, an
o
aO
q.
*6
90o
. Q
If
s >
O do
a
o
*
8o
Q
< ?
o 00
o
o C O
o
o
a!?
H
Australian Earthquakes to 1994
Magntudes: *405o
Figure 1. Australian earthquakes with magnitude exceeding M L4.0 since 1850 (Gibson, 19 94).
Earthquakes on reverse faults usually give a
high stress drop, where the seismic energy
comes from a small source volume. High stress
drop earthquakes radiate a greater than average
proportion of their energy in higher frequencies.
Higher frequency motion implies higher
accelerations for a given energy release or
earthquake magnitude, but not necessarily
fewer cycles, than for similar magnitude
earthquakes in, say, West Coast USA.
Compression giving reverse and thrust faul
produces surface uplift. Therefore, earthquakf
are most likely to occur in areas whei
mountains are developing, and less likely und0)
jjquiiocScn
Pm+mnca\ Mo
Jogontt daw (
CtMHdol , A4
> Iw^BeNan
2030
*0
40
Figure 14. Relationship between stress ratios causing liquefaction and (N,),# values for clean sa
magnitude 7.5 earthquakes. (Seed and De Alba, 19 86).
2030
< N * 0
Figure 15. Relationship between stress ratios causing liquefaction and (N,)^ values for silty sa
magnitude 7.5 earthquakes (Seed and De Alba, 1986).
(iii) For earthquakes of magnitude other
than 7.5, correct the values of Tav/a'0 by
the factors in Table 13.
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Table 13. Representative Number of Cycles and Corresponding Correction Factors (Seed and De
Alba, 1986).
Earthquake
Magnitude (M)
Number of
Representative Cycles
at 0.65 t
Factor to Correct
Abscissa of Curve in
Figures 14 and 15
8.5
26
0.89
7.5
15
1.00
6.75
10
1.13
6.0
5-6
1.32
5.25
2-3
1.5
It should be noted that the magnitude of
the earthquake has a significant
influence on whether liquefaction will
occur. It may be assumed that
liquefaction will not occur for
earthquakes of Magnitude 5 or less
(there are not sufficient cycles of
loading for small earthquakes).
However, where static liquefaction may
occur, even small earthquakes could
trigger failure. Morgenstem (1995)
discusses static liquefaction which is
likely to occur only in very
loose/poorly compacted, saturated soils
including dredged sand, mine
overburden dumps and mine tailings.
Salmon (1995) points out that when
earthquake loading is determined by a
probabilistic analysis of the history of
earthquakes in the seismotectonic zone
around the dam (as is done in
Australia), it is possible to determine
the contribution of different magnitude
earthquakes to the assessed ground
motion. Figure 16 gives an example
which shows that the major
contribution to the estimated ground
motion for a given PGA (in this case
the 1 in 1000 PG A) is from small
magnitude earthquakes near the dam.
Given the nonlinear relationship
implied in Tab le 13, and Figures 14 and
15, this has an important influence on
the assessment of the probability of
liquefaction, and should, if practicable,
be included in the assessment.
/
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50
GO
OS
I t
o
10
50
Suniat9 0
M A G N IT U D E C O N T R IB U IIO N S
-t-
Sums5 6
lb.
34567
M A C N I IU O E ( M )
oisiANcr coNfRinunnN^
Nnle ConUfbolKjftS /ro
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- 4
o
a.
c
o
2 3
- 0
O JomfoUowtki tf oL 09091
0 MurorrocM and KoboroiM UM 2)
A (ihihofo and K090 (Oil )
it Robvrlion (1902)
V MHehtll (1903)
O HoriNr > oL (1964)
ne '
bkul/lool
OO 002 003 01 0205 I
Mean Groin Sit*, O0 -mm
Figure 17. Variation of ratio with mean grain size (qc) measured in tsf 100 kPa) (Seed and De
Alba, 1986).
06
V 0'
Z 03
>
u
o
o.i
-11-
M> 7.3 orthquaku
% riots >33 213 slO 13
Ojolmm) 01 0.2 0.2302304 08
le/Ngo 3.3 41 4.4 4,4 4 33
_1_
X
4080120160200
Modifiad Con* P*n*tralion R*iilanc*. qe|-lf
240
Figure 18. Relationship between stress ratio causing liquefaction and cone tip resistance for sands and
silty sands (1 tsf100 kPa (Seed and D e Alba, 19 86).
There are several problems in applying the Seed
et al semi empirical approach:
(a) It has been developed for level or near
level ground conditions. For most dam
applications it can therefore only be
directly applied to assess whether
liquefaction would occur without the
dam.
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Seed and Harder (1990) recommend
that to allow for the static driving stress
due to the dam, a correction factor
should be applied to the calculated
(Tav/CT'0). This is calculated using
(xayCT'0)a^=(xav/a'0)a^ Ka
where Ka is a correction factor
determined from Figure 19. To
determine Ka, the relative density and
a (the ratio of static driving shear stress
on a horizontal plane to the initial
effective overburden stress, (tav/a'0) has
to be determined, eg. from finite
element analyses.
The correction only applies to soils
where Tav/a'0
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(c) In most natural soil deposits the SPT
values are variable (over that increase
which occurs with the increasing
overburden stresses). There are no
clear guidelines given by Seed et al as
to how this should be accounted for.
US SR (1989 (a)) discuss this issue and
present some useful practical points
which are recommended for use. These
include:
The results of each interval of each
drill hole, with regard to liquefaction
potential, should be prepared in a
table and should be presented on
geologic cross sections and profiles
to allow examination of the
frequency and continuity of those
intervals indicating liquefaction
susceptibility. From such a
presentation a judgement is drawn
as to whether or not the continuity
of potential liquefaction intervals
indicated is great enough to be of
concern.
If the deposit being sampled is
known to contain, or may contain
gravel, these coarse particles may
increase the blow count, implying a
more dense, less potentially
liquefiable soil. To check for this,
USBR (1989(a)) recommend
recording SPT blow counts for each
300mm of penetration, and
correcting for the erratic effects of
gravel (as a minimum, the three sets
of 150mm blow counts should be
checked to see whether this potential
irregularity is present). Where
gravel is extensive, shear wave
velocity methods should be used to
assess liquefaction potential.
X
}*
*x
V
X
X
X
5s .* *
X
t
X
35%
=0 ifFC
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It is important to note that a positive result from
the Seed method only indicates that a soil may
be susceptible to liquefaction. It does NOT
mean that just because (part of) the dam
foundation will liquefy, the dam will fail. What
should follow such an outcome, is an
assessment of the extent and continuity of
potentially liquefiable soil, and the residual
undrained strength of that soil, for use in post
liquefaction analysis.
Liao, Venziano and Whitman (1998) provide
some useful information which will help in the
application of liquefaction in a probabilistic
framework.
5.3.4 Shear Wave Velocity Methods for
Assessing Liquefaction Potential
Shear wave velocity is affected by many of the
variables which influence liquefaction, eg.
(relative) density, confining pressures, stress
history, geologic age. Hence, it has some use as
an indicator of potential for liquefaction.
The shear wave velocity may be obtained by
downhole, crosshole or surface to downhole
seismic methods, or by a seismic cone
penetration test (a modification of the
piezocone test) developed by Campanella et al
(1986).
As discussed above, some coarse grained
cohesionless materials (gravels, cobbles)
suspected of being potentially liquefiable
cannot be successfully sampled using SPT. If
crosshole shear wave velocity data have been
obtained on the materials, and these data
accurately represent the deposits in plan and
section, then they provide a viable means for
making a judgement on liquefaction potential.
USBR (1989) recommend that:
if shear wave velocities are >365m/s, the
deposits may be judged non liquefiable
if shear wave velocities are between 245 and
365m/s, the deposit may be considered
likely to be non liquefiable, but supporting
evidence should be obtained
if shear wave velocities are < 245m/s, the
deposit may be judged liquefiable.
In practice, this will leave many sites in the
"grey" zone.
Shear wave velocity may also be used to assess
liquefaction by relating peak ground
acceleration, and a history of performance of
sites in earthquakes. Bierschwale and Stokoe
(1984) and USNRC (1985) give details.
US N RC (1985) also gives details of a method
which assesses peak strains due to the
earthquake from the shear wave velocity, and
compares this to threshold strain. Brief details
of these methods are given in Fell et al (1992).
5.3.5 Determination of Residual Undrained
Strength
In recent years there has been quite a lot of
discussion of the post liquefaction condition.
This is usually discussed in terms of "residual
(undrained) strength", "field residual strength",
or "steady state undrained strength". Some of
these are expressed as plots of residual
undrained strength versus SPT 'N' value (eg.
Seed (19 87), Lo and Klohn (1990), Seed and
Harder (1990). These plots are developed from
backanalyses of liquefaction failures. Figure 22
is the plot presented by Finn (1993).
Finn (1993) indicates that lower bound
strengths are often used for these analyses,
although 33rd percentile values have sometimes
been used. In either case, the lower bound or 33
percentile gives very low (-zero) residual
undrained strengths at SPT less than about N =
6.
An alternative approach adopted by a numb er of
authors including Ishihara (1994) and Finn
(1993), is to relate the normalised residual
undrained strength Su/ct'vo (where SU5 = residual
undrained strength and ct'Vo effective vertical
stress) to either SPT (N,)^ values (N values
corrected to 100 kPa effective stress, 60%
energy ratio hammer) or to cone penetration
resistance qc.
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2000
I60C
H
O
Z
tu
cr
e
1.3 and maximum
compressive stress
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Figure 34 Sequence of Analysis for Operating Basis Earthquake
(Guthrie 1986)
The preceding method could probably be used
for arch and buttress dams as well. However,
consideration would have to be given to the
appropriateness of the damping ratios and
corresponding limiting tensile strengths used.
(c) Non-linear Dynamic Analysis
A non-linear dynamic analysis of a concrete
dam is a complex analysis and would normally
be undertaken by specialist numerical analysts
experienced in such work. It would normally
be done only for major dams where the cost of
the new dam or the cost of remedial works for
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an existing dam is sufficient to justify the
greater expense of this type of analysis.
Provided peak tensile stresses (static +
dynamic) do not cause cracking in the dam,
then a linear analysis will suffice. When
cracking occurs, stresses will re-distribute.
Therefore, when cracking is expected or there
are pre-existing cracks (e.g. open vertical
contraction joints) and the dam is essentially
3D m nature then a non-linear analysis should
be done.
There are three main ways in which cracks can
be modelled:
as distributed (smeared) cracks (zero elastic
modulus in direction perpendicular to
cracks)
as discrete cracks at finite element
interfaces
using fracture mechanics.
The first of the alternatives is computationally
the simplest. Further details on cracking are
given by Zienkiewicz, Valliappan and King
(1968), Rashid (1968), Mohraz, Schnobrich
and Gomez (1970), Darwin and Pecknold
(1978), Phillips and Zienkiewicz (1976),
Bazant and Cedolin (1979), Bazant and Ob
(1979), Argyris, Krempl and William (1977),
Gerstle (1981), Kotsovos and Newman (1978),
William and Warnke (1975), Cedolin, Crutzon
and Dei Poli (1977), Bicanic and Zienkiewicz
(1983), Zienkiewicz, Fejzo and Bicanic
(1983), Zienkiewicz, Hinton, Bicanic and
Fejze (1980), Pande and Shen (1982), Pal
(1974) and Chapuis, Rebora and Zimmermann
(1985).
A non-linear analysis will by its nature,
require a time-history analysis. Consideration
will have to be given to:
the way the compressibility of the storage
water is modelled
the way seepage pressures particularly
those due to water penetrating cracked
zones, are modelled.
The non-linear analysis of concrete dams is
still a developing and specialised field. Other
relevant papers include Waggoner, Plizzari
and Saouma (1993), Gao Lin, Jing Zhou and
Chuiyi Fan (1993), Greeves and Taylor
(1992), Cervera, Oliver and Galindo (1992),
Jing Zhou and Gao Lin (1992), Clough and
Ghanaat (1993), Fenves and Mojtahedi (1993).
7.3.4 Analysis of Permanent Deformations
In some concrete gravity dams subjected to
say the maximum credible earthquake, it may
be permissible for the dam to slide on its base
or within the foundations and be permanently
deformed after the earthquake. This of course
assumes that during or after the deforming
process, the security of the storage is not
preudiced allowing for the potentially
increased uplift pressures and lower strength
which may apply. Researchers such as Chopra
and Z hang (1991), Leger and Katsouli (1989)
and Danay and Adeghe (1993) discuss
calculations which indicate that typical
permanent displacements for large concrete
gravity dams subjected to earthquakes having
peak ground accelerations the order of 0.5g
can range from tens of centimetres to more
than half a metre. However, some dams with
suitable foundations would be able to
withstand small displacements. Dams relying
on post-tensioned ground anchors or drain
holes for normal load static stability, might not
be able to withstand these sort of movements
if the displacement was sufficient to shear the
anchors. However it may be acceptable to
have the anchors sheared for a low probability
earthquake provided the dam was stable under
the post earthquake load case.
The type of analysis required to compute
permanent deformations is similar to that
carried out for embankment dams i.e. Makdisi
and Seed (1978). The dam is considered as a
block with a limiting sliding strength along its
foundations. The dam is subjected to a time
varying input of acceleration. When the
acceleration is greater than the limiting
acceleration (the acceleration causing inertia
forces which are greater than the sliding
strength of the foundations) the dam will move
on its foundations. The parts of the
accelerogram greater than the limiting
acceleration are double integrated to obtain
cumulative displacements.
The type of analysis just described is a
requirement of the US Corps of Engineers
method for dynamic analysis of concrete
gravity dams when the sliding safety factor is
less than one. The sliding analysis is carried
out for horizontal planes through the dam
where there is cracking. A crack is assumed
through the dam with suitable slip elements
along the crack interface.
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In their study of base sliding, Chavez and
Fenves (1993) found that sliding was not
likely to occur if the storage is less than half
full. Other conclusions were :
vertical ground motion has almost no effect
on the sliding displacement (it slightly
increased the maximum stresses). This is
partly because the vertical and horizontal
ground motion do not necessarily coincide.
the assumption of rigid foundation rock can
significantly overestimate the amount of
base sliding
At the present stage of development of the
computation of permanent deformations and
the determination of suitable maximum
displacements there is still much work to be
done especially regarding local conditions.
These guidelines therefore counsel that
considerable care be used if a dam is to be
allowed a permanent deformation following an
earthquake. Adequate sliding and overturning
stability must exist after the earthquake using
foundation strengths appropriate to the
displacement (usually the residual strength)
and uplift appropriate to the displaced
condition, allowing for opening of joints and
bedding, and for reduced (or no) drainage
capacity.
7.4 Design Earthquake and
Hydrodynamic Loads
7.4.1 Earthquake Parameters
As discussed elsewhere in these guidelines,
dynamic analysis can be carried out in the
frequency domain or the time domain. For the
former, response spectra are required while for
the latter, accelerograms are required.
(a) Response Spectra
A response spectrum shows the extent to
which any single degree of freedom structure
with an assumed level of damping would
respond to particular earthquakes.
Knowing the natural frequencies of vibration
and the corresponding mode shape for a
structure, the spectral accelerations
corresponding to particular natural frequencies
and damping ratios can be converted to inertial
loads.
The response spectrum used in the analysis of
a dam should be site specific and relate to the
peak ground accelerations examined. The
response spectrum should also reflect the
frequency mix and duration of the design
earthquakes. It will therefore be derived from
a number of earthquakes having various
epicentral distances from the site and
consequently, different acceleration
attenuation functions. The response spectra
should be obtained from a seismologist as part
of the assessment of seismicity of the dam site.
(b) Accelerograms
Where a time-history analysis is to be done, at
least three different accelerograms appropriate
to the dam site and for a particular peak
ground acceleration, should be used. These
accelerograms may be recorded accelerograms
which are suitably scaled (accounting for
change in frequency mix and phase with
change in peak ground acceleration) or
synthetic accelerograms which fit the response
spectra for the site. Care should be taken in
selecting accelerograms which are similar to
Australian earthquake conditions, and advice
should be obtained from a seismologist.
7.4.2 Hydrodynamic Pressures
Any movement of the dam and foundation will
cause movement in the water of the storage
and in turn, the pressures generated by the
water will impose forces on the dam.
Engineers have traditionally used
hydrodynamic pressures derived by
Westergaard (1933). These pressures are
commonly converted into equivalent lumped
'virtual' masses which are attached to the dam.
Westergaard's pressure distribution assumes
that the water in the storage is incompressible
and that the dam and its foundations are rigid.
However, this is not always so. In high
gravity dams and slender arch dams especially,
where the dam is flexible, there can be
considerable interaction or coupling between
the dam and the storage.
Considerable work on the interaction of
gravity dams and their storages has been done
by Chopra and his fellow researchers. Details
of the work are given in Chopra (1967),
Chakrabarti and Chopra (1974), Chopra,
Chakrabarti and Gupta (1980), Chopra and
Gupta (1981). Other relevant papers include
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Clough and Chang (1980), Dungar (1978),
Hall (1988), Zienkiewicz, Paul and Hinton
(1983) and Tsai and Lee (1989).
Besides the lumped, 'virtual' mass approach,
the interaction of a dam with the water in its
storage can be determined by treating the
water as a 'solid' having zero shear modulus
but retaining compressibility. This approach,
although simple in principle, has a number of
numerical problems. An alternative approach
is therefore preferable where from the
beginning, the shear components of stress in
the fluid are neglected. This latter approach
includes the effect of water compressibility. It
also will allow the limiting case of
incompressible water to be considered.
Accounting for water compressibility and
coupling of a dam with the water in its storage
adds considerable computational effort to
determining the effects of earthquake loading
on a dam. Neglecting coupling and water
compressibility in the simpler "added mass"
approach is not considered significant for
excitations at frequencies below the natural
frequency of the reservoir.
An estimate for the fundamental frequency of
a reservoir can be obtained from:
f =__
w4H
eff
where: fw = the fundamental frequency
C = the compression wave speed
in water (l,439m/s)
Heff = an effective depth of the
reservoir.
The above relationship was obtained from
Duron, Ostrom and Aagaard (1994) and
applies strictly to an infinite reservoir of
constant cross section. Duron and Hall (1988)
indicate that if the ratio of to the
fundamental frequency of the dam-foundation
alone (fj) is near unity, water compressibility
will have a significant effect. For ratios of fw
to fj much greater than one (e.g. >1.5)
incompressible fluid behaviour can be
assumed.
7.4. i Uplift/Seepage Pressures
The USBR (1977) considers that the uplift
pressure within the crack is zero while ICOLD
(1986) assumes full headwater pressure but
recognises the need for further research.
Guthrie (1986) uses the pre-crack uplift
pressure diagram.
These guidelines recommend that for the
duration of the earthquake, the pre-earthquake
uplift pressure distribution is used for the
stability analysis. However for the post
earthquake analysis, consideration should be
given to the amount of cracking and the post-
earthquake efficiency of the dam's drainage
system. In the post-earthquake situation, full
headwater pressure is assumed to exist in a
crack at least as far as the line of drains. If the
drains have sufficient capacity and they have
not been disrupted by sliding of the dam, then,
if the crack extends, past the drains, a
significant reduction in uplift pressure should
be considered. Typically, the pressure at the
line of drains in this case might be the
tailwater pressure plus 50% of the difference
between headwater and tailwater pressures.
The pre-earthquake uplift pressure distribution
might have had a 67% reduction. The lesser
reduction for the post-earthquake case allows
for the greater amount of drainage with which
the drains would have to cope.
7.5 Design Criteria
7.5.1 General Approach
The working group has had two major
difficulties in preparing suitable design criteria
for concrete gravity dams.
(1) The current ANCOLD (1991)
guidelines for design criteria for concrete
gravity dams are based on a limit state
approach with partial factors of safety. This
method has proven to be difficult to use on
dams subject to significant earthquake loads.
The ANCOLD (1991) guideline is under
review to address these problems. In the
interim, it is recommended that the design
loadings and acceptance criteria described
herein are used. These are based largely on
the BC Hydro (1995) guidelines.
(2) Existing guidelines for design of
concrete gravity dams are not simply applied
to a risk based approach. As a result it has not
been practicable to develop these guidelines to
directly apply to a risk based approach, and
they are given in terms of a deterministic
method using OBE and MDE. For
completeness, flood and static load case are
also listed. Those wishing to use a risk based
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approach may do so, taking account of the
general intention of the load cases detailed
herein.
7.5.2 Loads and Load Cases
The loads considered in the assessment of
concrete gravity dams and their foundations
should include the following:
Dead loads of permanent structures,
equipment and foundation rock (D).
Water load due to maximum normal
headwater level combined with the most
critical concurrent tailwater level (H).
Water load due to maximum flood
headwater level based on the Inflow Design
Flood (IDF) with corresponding tailwater
levels (H).
Foundation uplift (U), both at the concrete-
rock contact and at critical discontinuities
within the foundation.
Static and dynamic thrust created by an ice
sheet, for reservoirs subject to freezing (I).
Vertical and horizontal loading due to rock
or soil backfill (both natural or engineered),
including potential effects of liquefaction
and loads from silt deposited against the
dam (S).
Load due to Operating Basis Earthquake
(OBE) Q'
Load due to Maximum Design Earthquake
(MDE ) (Q).
Determination of the loads should take into
account the actual field conditions and
instrumentation records.
Foundation uplift assumptions should reflect
the stress state and condition of the dam and
foundation. Disruption of the dam and/or
foundation condition due to an earthquake
should be recognised in assessing the uplift
assumptions for the post-earthquake case.
The dam and foundation should be assessed
for the following load cases:
(a) Usual Load Case
permanent and operating loads should be
considered for both summer and winter
conditions including self-weight, ice (where
applicable), silt, earth pressure, and the
maximum normal reservoir level with
appropriate uplift pressures and tailwater level.
(D + H +1 + S + U)
(b) Unusual (Flood) Load Case
Permanent and operating loads of the Usual
Load Case, except for ice loading, should be
considered in conjunction with reservoir and
tailwater levels and uplift resulting from the
passage of the IDF.
(D + H' + S + UF)
where subscript "F" refers to the flood case.
The potential should also be assessed for
reservoir levels higher than would result from
passage of the IDF, such as those due to
operating failures or other unusual conditions.
The effects of ice loads should not be
considered simultaneously with flood
conditions.
(c) Unusual (Earthquake) Load Case
Permanent and operating loads of the usual
load case except for ice loading, should be
considered in conjunction with earthquake
loading associated with the Operating Basis
Earthquake (OBE). The effect of ice loads
should not be considered simultaneously with
OBE earthquake conditions.
The analysis should be carried out for the dam
empty case.
(d) Extreme Load Case
Permanent and operating loads of the Usual
Load Case should be considered in
conunction with seismic loads of the
Maximum Design Earthquake (MDE).
(D + H + S + Q + U)
The effects of ice should be given special
consideration, recognising the high uncertainty
associated with ice loading on earthquake
loading, and its effects on the dam.
(e) Other Load Cases
Where earthquake-induced cracking at the
concrete-rock interface or any weak section is
identified, a stability analysis should be
carried out to assess whether the dam in its
post-earthquake condition is capable of
resisting loads of the Usual Load Case.
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(D + H + S + UpQ)
where subscript "pg" refers to the post-
earthquake case.
Concurrent ice loading (with the post-
earthquake condition) may be considered in
areas where appropriate.
A landslide generated wave case should be
considered where existing or potential
landslides, which may affect the reservoir,
have been identified. Combinations with other
loads would be site specific.
An inoperative drain case assuming plugged
drains may be assessed and could be
considered as an Unusual Load Case.
7.5.3 Acceptance Criteria
All kinematically feasible failure modes,
analysed by the single-slice, rigid-body, force
equilibrium method, should satisfy the
acceptance criteria shown in Table 18.
Table 18
Stability Index Acceptance Criteria
Load
Sliding Factor
(Note 1)
Position of Resultant Force
(Note 2)
Minimum Compressive
Stress Factor
(Note 3)
Usual
1.5-2.0
Mid-third of surface
No tension
4.0
Unusual (Flood)
and
Unusual (OBE )
1.3-1.5
Mid-half of surface
One-quarter tension
2.7
Extreme (MDE)
1.1 - 1.3
Within surface
1.3
Post-earthquake
1.2-1.4
Mid-half of surface
One-quarter tension
2.7
Notes: 1. Sliding Factor (Frictional Analysis) = Resisting Forces
Applied Force
Lower values of the range apply where the geology and the strength parameters are
reasonably well known.
2. Vector summation of all forces, including uplift, acting on the analysis surface
3. Compressive Stress Factor = Unconfined Compressive Strength
Compressive Stress Normal to Surface
To be considered primarily for massive but low strength rock and weak deteriorated
concrete.
7.5.4 Post Earthquake Stability
If a dam is likely to be severely damaged after
being subjected to the MDE, considerable time
may elapse before the dam can be repaired or
the storage lowered. Consequently, all parts of
the dam will need to remain stable after an
extreme earthquake event. The stability of the
dam should therefore be checked for static
loading conditions. The assumed uplift
pressure distribution should be as discussed in
sub-section 7.4.3, i.e. full headwater pressure
within cracks emanating from the upstream
face.
7.5.5 Foundation Stability
Where there is the possibility of a sliding
failure along faults, shears and/or joints, the
stability of the foundations should be
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examined. This examination should be made
for both the earthquake and post-earthquake
cases. The dam itself should be examined for
local overstressing due to foundation
deficiencies.
Sliding failure is especially likely when
discontinuities and/or horizontal or sub-
horizontal seams close to the foundation
surface contain clay, or have been previously
sheared eg. bedding surface shears due to
stress relief, folding, or associated with faults.
7.6 Dynamic Material Properties
7.6.1 Concrete
The compressive and tensile strengths of
concrete increase with increased rate of
loading. The dynamic compressive and tensile
strengths of concrete can therefore be expected
to be greater than the static strengths. As
dynamic compressive stresses are rarely of
concern, the allowable compressive stress for
static loading can be used also for dynamic
loading.
Raphael (1984) states that the apparent tensile
strength of concrete under seismic loading
which should be used with linear finite
element analyses is given by:
fr =065 fc 2P
where fc is the concrete compressive strength
in MPa and fr is the apparent seismic tensile
strength in MPa. Values given by this formula
are some 50% greater than the apparent tensile
strength for static loading. Raphael suggests
that fr is twice the splitting strength of the
concrete under static loading.
C lough and Ghanaat (19 93) suggest that the
apparent dynamic tensile strength is about
25% greater than the measured static value
which gives apparent tensile strength about
20% of the standard compressive strength.
They further suggest that there may be a 15 to
20% loss of strength across lift joints. These
figures may be even lower for poorly
constructed or defective lift surfaces,
fiowever, the peak dynamic tensile stresses
only exist during a fraction of a response
cycle. Even though these peak stresses may
greatly exceed the tensile strength of the
concrete, any cracking that might be initiated
will not have time to fully develop. It is well
recognised that a single spike of excessive
localised tension should not be taken to
represent dam failure.
In consideration of the above however, these
guidelines recommend that for sound lift
surfaces, the apparent tensile strength to be
used is 16% of the standard compressive
strength.
For dynamic modulus of elasticity, Clough and
Ghanaat (1993) suggest a value 25% greater
than the static value and these guidelines
recommend this be adopted. For existing
dams, the elastic modulus of the concrete mass
may be determined using geophysical means
(e.g. derived from measured shear wave
velocity). Values obtained should be
compared with static and dynamic small
sample laboratory test values for credibility.
7.6.2 Rock
In most cases the stability of the dam will be
controlled by sliding in or on the dam rock
foundation. To carry out static and dynamic
analyses, it will be necessary to:
assess and map surface exposure,
excavations and drill core to define the
geology of the site. In particular the 3-
dimensional orientation, continuity and
detailed nature of bedding, joints, and other
features such as shears are required.
the shear strength of the foundation rock
should be determined using appropriate
rock mechanics techniques, such as those
described in Hoek (1983, 1990, 1994),
Hoek and Brown (1980), Patton (1966),
Barton and Choubey (1977), Barton and
Bandis (1991). These require an
assessment of the orientation, spacing,
continuity, shape and roughness of
discontinuities in the rock (e.g. joints), and
the strength of the rock substance as this
varies with confining stress. Care should
be taken in applying these techniques, to
account for the presence of continuous,
adversely oriented low strength surfaces
such as bedding surface shears, faults or
shears, and to take account of the
mechanisms of failure.
The Hoek, and Hoek and Brown methods give
strengths for "undisturbed rock" and
"disturbed rock". The latter should generally
be adopted unless advice from an experienced
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rock mechanics specialist indicates otherwise,
because of the uncertainty as to the accuracy
of the Hoek methods, and since the
undisturbed rock strengths apply to confined
conditions such as underground openings,
where dilation or shearing causes increases in
normal stress.
It will be noted these stress methods give non
linear failure envelopes, with high friction
angle and low cohesion at the normal stresses
usually applicable to dams. They also allow
estimation of the modulus of the rock mass
(Hoek, 1994). The dynamic modulus may be
higher than the static modulus as discussed in
Clough and Ghanaat (1993) and Scott and Von
Thun (1993) and may, as for concrete, be
obtained by geophysical means, or by relation
to the static modulus.
A dam's foundations will normally contain
joints, shears, and bedding. Consequently, it
will not be possible to transmit tensile stress
within the foundations and the allowable
tensile strength for the foundations will
therefore normally be assumed to be zero.
However, if extensive site investigation and
strength testing is able to prove that the
foundations for a particular dam site are
capable of transferring tensile stresses, then
the tensile strength of the rock may be
included.
Where foundation rock strength becomes
critical, as they often will, advice should be
obtained from a person expert in rock
mechanics.
8. APPURTENANT
STRUCTURES
8.1 Introduction
A number of subsidiary structures associated
with a dam are essential for the dams
operation. Consequently damage to or
destruction of these appurtenant structures
would be prejudicial to the dam's safety. An
important facility at a dam is one that allows
water to be released in a controlled manner. If
therie has been an earthquake and the dam is
damaged to the extent that the dam is not
serviceable then it may be necessary to lower
the storage so that remedial works can be
undertaken. It will therefore be necessary that
not only the outlet structures and their gates
and valves remain serviceable but also that
there is proper access to these structures.
Bridges and roads may need to remain in a
sound state after an earthquake depending on
their importance.
Generally, appurtenant structures should be
such that:
they maintain their normal operating
condition after an operating basis
earthquake
they are not damaged to an extent where
they could allow sudden or uncontrolled
loss of water from the storage for a more
extreme earthquake up to the maximum
design earthquake.
8.2 Performance Requirements
This sub-section gives the performance criteria
for the operating basis earthquake (OBE) and
the maximum design earthquake (MDE) which
could be the maximum credible earthquake
(MCE). Performance requirements are given
for a number of appurtenant structures
including intake towers, outlet conduits, outlet
works, spillway gates, spillway piers, spillway
gate hoist piers, access bridges and piers,
access roads.
Intake Tower
OBE: Static and dynamic loads to
induce maximum concrete and
steel reinforcement stresses
which satisfy AS3600
(Concrete Structures Code)
(i.e. limited amount of
reinforcement yielding) and
the tower and its base remain
stable.
MDE: Significant amount of
reinforcement can yield
horizontal reinforcement
designed to prevent vertical
reinforcement from buckling
and to contain concrete when
it is in compression (i.e.
concrete contained between
inner and outer layers of
vertical reinforcement will not
spall away).
Outlet Conduit
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O B E :
MDE:
Outlet Works
O B E :
Static and dynamic loads to
induce maximum concrete and
steel reinforcement stresses
which satisfy AS3600
(Concrete Structures Code).
Dynamic loads to include
those induced from the
earthquake effect on the
overlying dam.
Conduit not to collapse or
rupture - Collapse could lead
to undermining and
subsequent failure of an
overlying embankment.
Rupture could cause piping or
destabilise an embankment by
a marked increase in pore
pressure.
All valves to maintain their
normal operating capabilities.
MDE: Emergency closure and
regulating valves (especially
low level release valves) to
maintain operating capability -
storage may have to be
quickly lowered if parts of the
dam are damaged and need
remedial works or relief of
hydrostatic loads.
Spillway Gates
OBE: Gates to maintain normal
operating capability.
MDE: Gates retaining permanent
storage at the time of an MDE
should not fail to the extent
where water from the storage
is released in an uncontrolled
manner. MDE should not
cause the gates to distort to an
extent that they cannot be
opened or closed.
S pillway Piers
O B E :
Carry out appropriate dynamic
analysis for the spillway piers
for earthquake loading in the
upstream/downstream and
transverse directions.
Combined static and dynamic
loads should satisfy ASS 600
(Concrete Structures Code).
Earthquake loads from the
orthogonal directions may be
combined on a square root of
the sum of the squares basis.
MDE: Carry out dynamic analysis as
for the OBE. Combined static
and dynamic loads may cause
cracking but piers must
remain stable for overturning
and sliding. Piers should not
be permanently displaced to
the extent where the spillway
gates become jammed.
Spillway Gate Hoist Piers
OBE: Carry out appropriate dynamic
analysis similar to that for
spillway piers. Combined
static and dynamic loads
should satisfy AS3600
(Concrete Structures Code).
MDE: If the gates can be operated
from an alternative position
(possibly, in a less efficient
manner), then the spillway
gate hoist piers (and hoist
bridge) can be allowed to fail.
If the continuing operation of
the gates depends on the
continued viability of the hoist
bridge piers (and hoist bridge)
then carry out an appropriate
dynamic analysis similar to
that for the spillway piers.
Combined static and dynamic
loads may cause cracking but
the piers must remain stable
for overturning and sliding.
Note: theappropriate
dynamic analysis
required for spillway
piers and spillway
hoist piers may be part
of an overall dynamic
analysis for a concrete
dam.
Access Bridge arid Piers
OBE: Carry out appropriate dynamic
/ of the pier and bridge
'The anaysis may be
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part of an overall dynamic
analysis for an intake tower.
The pier and bridge system
should be examined for an
earthquake in the direction of
the bridge access and
perpendicular to the bridge
access. Earthquake loads may
be combined on a square root of
the sum of the squares basis.
Combined static and dynamic
loads should satisfy the
appropriate Structures Code).
MDE: If there are alternative means
of access then the access bridge
can be allowed to fail. If there
are no alternative means of
access then the bridge and piers
should remain capable of
carrying their design loads.
Access Roads
OBE: Access roads to the dam and
its appurtenant works should
remain passable immediately
after the OBE. Therefore any
likely land slip areas along the
access roads should be checked
for stability. There should be no
land slips either during or after
an OBE.
MDE: Access roads to the dam and
its appurtenant works may
become impassable during or
immediately after an MDE.
However, they should be easily
cleared. Therefore, while there
may be land slips onto the
roads, the roads themselves,
should not be allowed to
collapse where there is no easy,
alternative access route.
8.3 Intake Towers
8.3.1 Analysis
The analysis method for intake towers is
described here in general terms only. A more
complete description of the method on which
the following general description is based, is
given by Chopra and Goyal (199 1) and Goyal
and Chopra (1989).
The method presented here is based on a
simplified dynamic analysis in the frequency
domain using a suitable response spectrum. It
uses an added mass representation of
hydrodynamic effects due to surrounding
(outside) water and contained (inside) water
(in the case of wet towers). In addition, it
includes the effects of tower/foundation
interaction.
The steps of the method are:
(i) Select suitable response spectrum.
(ii) Compute the added hydrodynamic
mass of water using Goyal and Chopra
(1989).
(in) Determine the structural properties of
the tower
mass per unit height
flexural stiffness
modal damping ratios.
(iv) Compute natural periods and mode
shapes for the first two modes of
vibration.
(v) Determine the spectral accelerations
for the first two modes of vibration
from the response spectrum.
(vi) Compute the generalised mass and
generalised excitation terms using the
mass distribution and the mode
shapes.
(vii) Compute the inertia forces using the
spectral accelerations, generalised
mass and excitation terms, mass
distribution and mode shapes.
(viii) Add the inertia loads from the first
two vibration modes on a square root
of the sum of the squares basis.
(ix) Compute bending moments and shear
forces from (viii).
(x) Design tower according to AS3600
(Concrete Structures Code).
8.3.2 Design Criteria
As discussed in Sub-section 8.2, an intake
tower is designed elastically for the OBE.
Generally, it will not be necessary for the
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tower to behave elastically during the MDE.
In order to ensure that the tower can survive
intense ground shaking due to the MDE with
limited damage, it should possess a ductility
capacity greater than the ductility
requirements imposed by the ground motion.
A suitable method for this is described in
Chopra and Liaw (1975) where a ductility
factor of two is recommended (ratio of
maximum permissible displacement to the
yield displacement).
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R E F E R E N C E S
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